Deflection of Reinf. Masonry Wall backing up brick 5

[jechols]

I am currently designing a 32'-0" tall, 12" block wall supporting long span roof joists. I have determined for my loading conditions that #5 at 24" o.c. is adequate for strentgh. When I presented this to the boss he thought deflection might be a problem since the wall is backing up brick. I usually do not perform a deflection check for out-of-plane loading on a masonry wall (IBC 2003/ASD). I have determined the theoretical defl. to be about 1.5" and L/600 equals 0.64". Any thoughts? To meet the L/600 I would have to grout the wall solid.

Add'l info:
DL 1650 plf
LL 1100 plf
wind 16 psf
sds 0.32
6' parapet

Thanks in advance for your comments!

 

[csd72]

With that sort of axial load, you may be able to justify the bottom as fixed (for deflection) which will help bring the deflection down.

csd

 

[jechols]

csd72,

thanks for the input. I guess my real question is simply do you think a reinforced masonry wall backing up brick veneer has to meet the L/600 criteria or just the 0.007 H limit stated in section 3.2.5.6 of the ACI 530-02?

 

[lkjh345]

What are you using for the eccentricity of your axial load?

If you are using a 6" deep ( parrellel to direction of joist span) bearing plate, with edge set back 1/2" from edge of CMU, (Per section 104.4 of SJI Standard Spe) your eccentricity would be 2.31 inches.

 

[csd72]

Depends on which building code you are using.

Also,are you multiplying the wind loads by 0.7 for serviceability?

 

[jechols]

lkjh345,

Based on the 1/2" from the edge as you stated and the diagram (on page 43 of the Vulcraft manual) of the top chord bearing of LH / DLH joist which shows the resultant reaction 2" from the end of the joist I calculated the eccentricity to be 1-5/16". I used 1.5" for my design. Would you agree or am I missing something? On a side note, do you put a minimum eccentricity on a wall (0.1xwall thickness). I only see that requirement for a column.

Thanks,

j

 

[lkjh345]

jechols:

You are correct. Using the diagram on P43 of Vulcrafts manual (with center of loading 2" from end of joist) you do get about 1.5 inches of eccentricity.

Personally, I usually use 1/6 of the wall thickness as minimum eccentricity. I don't remember where I picked that up at, so it may or may not have any validity behind it.

A somewhat outdated book by James Amerhein (published by Masonry Institute of America ) published in 1994 recommends using .007h as maximum lateral deflection of a wall to ensure structural integrity, but sort of criptically warns that this will not prevent brick cracking or water penatration.

 

[jechols]

lkjh345,

That makes sense. I am going to try to treat the wall fixed at the base as suggested by CSD72, multiply wind load by 0.7 (thanks CSD72), and increase bar size to reduce Mcr. Maybe by adjusting those 3 parameters I can at least get close to L/600.

Since technically I need a relief angle anyway, another option would be to install the angle at mid-height of the wall which would counteract worst case condition while also elminating L/600 requirement.

Thanks again to all who responded.

 

[JAE]

csd72 - you might provide a reference for that 0.7 factor. I'm not sure where that comes from.

 

[concretemasonry]

jechols -

The concept of a relief angle near mid height is outstanding!

Without the relief angle you may end up with an uncontrolled crack near that location IF you did have excessive deflection.

As usual, just make sure you flash it properly to shed moisture from the CMU wall.

This also solves the conflict between the dissimilar long term expansion of the brick in comparison to creep/shrinkage of the concrete masonry or any steel or concrete in the structure. A brick veneer wall 32' high without relief is stretching it as you recognize.

Dick

 

[Lion06]

JAE-
I believe the 0.7 factor is a result of the strength wind loads using 50-year wind speeds, but serviceability only needs to be designed for 10-year wind speeds. This is discussed very well in the AISC Design Guide #3 (I think) - The one about serviceability of steel buildings.
For non-hurricane zones, the factor for 50-year to 10-year is around 0.84. 0.84^2 is about 0.71, rounded down to 0.7.

 

[UcfSE]

JAE, that is in the footnotes for the deflection table in chapter 16 of the IBC and state codes based on the IBC. Not being at work, I can't give the table number right now.

I don't suggest you treat the wall as fixed. If you must have some restraint, I would try it as partially restrained at best.

 

[miecz]

UcfSE-

If the wall is dowelled to the foundation, why not consider it fixed at the base?

 

[Lion06]

jmeic-
It might be fine to treat the wall as fixed, but (Jechols) then be sure to go back and redesign the footing that way, because it probably wasn't designed as fixed at teh base originally. It will have to get bigger.

 

[miecz]

StructuralEIT

Well, maybe it should then. It the wall connection acts like a rigid joint, then the moments are transferred to the foundation. Ignoring them would underestimate the soil bearing pressure.

 

[csd72]

JAE,

The reference to the 0.7 factor is: 2003 IBC, Table 1604.3 footnote f.

csd

 

[JAE]

Yes - got it. I remmember that now from a past seminar and I have to say I rarely use it (I guess I'm just a bit conservative).

But thanks for the refresher folks - appreciate all of you that participate here. Keeps us all sharp(er).

 

[UcfSE]

There is a large difference between zero rotation and doweled into the footing. Even if the footing is desgned that way strength-wise, the soil will still deform causing the footing to rotate. What I'm getting at is just waving your arms and assuming the wall is fixed is bad. If you can show that the additional rotation and deformation results in acceptable deflections in the wall, the topic of concern, then maybe you have something. Even a directly-welded steel moment connection doesn't achieve zero rotation, and I wouldn't expect a cmu wall to perform nearly as well.

 

[miecz]

UcfSE-

True, the bottom of the wall does not have zero rotation. But, as you say, many connections are designed as fixed that aren't absolutely rigid, but are close enough. Correct me if I'm wrong, but I believe it's common practice to design spread footings as fixed supports without justifying the fixety assumption by analyzing the footing rotation. In fact, we had a recent thread where the flexibility of a footing/base plate/anchor bolt connection was left unanswered.

 

[UcfSE]

It sounds like someone needs to do some research then, if a question was left unanswered. This is not overly difficult given the internet and the nearest university library, not to mention all the education we've had to get to this point.

Common practice does not equal correct, it equals common. When designing a spread footing with a fixed base for instance, it may be up to our judgment to say that the extra deflection we get would not push us over the limit, or to put in some extra capacity to account for this. That is up to us as the engineer. The correct procedure is to account for those things. If judgment dictates not to, then don't. In order to know if it is close enough, you need to find the correct answer, compare wherer you are compared to the correct answer and make the judgment yourself. My judgment tells me this is not a good way to go in the case pertaining to the OP's question. Yours may be different. That is the way of things. The OP will have to make up his or her own mind and go from there.