Efficient Moment Frame Design

[cfox142]

Assuming you meet any prescriptive code requirements and connection design requirements etc how do you optimize a moment frame? It seems to make sense to me that you would want the moment of inertia of the beam and column to be roughly equal except in high seismic areas where you need strong column weak beam design. I am looking at 2 story building currently though and with that design my roof beams are stressed around 30% and my second floor beams are around 75% (deflection controls). It seems like I would want the stiffness to be similar for the entire structure. So I'm curious how others perform this task and optimize the structure. I am looking more for rules of thumb from experience than code requirements here.

Thanks,

 

[BUGGAR]

I go for the detailing. That's where the costs are. The sizes are almost secondary.

 

[KootK]

I agree with Buggar's approach 100%. That said, there is arcane theoretical background available out there for those so inclined:

Gravity Loads: Link
Lateral Loads: Link

The material weight sweet spot depends on the ratio of the beam to column span, which makes sense.

My, in practice, strategies basically amount to:

1) I like my columns beefy enough that they don't require continuity plates or doubler plates.

2) I like my columns to be traditional column sections so nothing bigger than W14 if possible. It's tough to LTB brace a W24 column intermittently.

3) I usually only change my columns every two stories, assuming that the columns will be erected in two story lifts.

4) I avoid getting anywhere close to creating a weak story situation with tall first floors.




I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[DETstru]

I'm with BUGGAR and KootK.

- Try limiting the quantity of moment connections. Even if your frames are heavier, the connections are very pricy. Try limiting the number of frame beams/columns to minimize cost.

- Turn your column strong axis in the direction of the higher unbalanced moment.

- Try making frame beams connected to a column all the same depth.

- Either design the connections to the demand required or give the fabricator the loads. Don't force excessive connection design by saying "design to full capacity..."

- Check to see if you can fix the bases of your columns. The foundations will be impacted but it may be minor. This will reduce deflection and weight.

 

[MrHershey]

Going through this right now, so will throw it out there. Keep your spans down in high seismic regions. NEHRP recommends 20-30 ft for economical special concrete moment frames, don't know what steel recommendations are but would imagine you'd similarly want to keep aspect ratios reasonable. If you've got large enough spans that gravity loading is the main force driving your beam sizes, your column size and/or reinforcing can start getting ridiculous quickly just to keep up with the beams. Pay special attention to this at roof levels where you only have one column below to offset beam strengths (as opposed to one above plus one below at normal levels).

 

[KootK]

@MH: you may find it useful to know that it's generally okay to violate strong column / weak beam at the roof level connections.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[MrHershey]

KootK, do you have a reference for that? I've seen that for steel but haven't seen much for concrete other than if you can keep your factored axial load under Agf'c/10.

 

[KootK]

I know that I first saw it in Paulay & Priestley. I'll dig a little deeper when I get back from the Bahamas.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[MrHershey]

Actually dug and found we have Paulay and Priestley in office and found what you're referencing, thanks.

In looking into it further, it intuitively makes sense. Wouldn't think it'd make a difference at roof level that your plastic hinge moves from beams to top of column. You're not creating a story mechanism there and we'd actually still be meeting SCWB for the joint below the roof level since you get to share moment to two columns (though P&P says a hinge there is usually fine too). Only thing I can think of is you'd cut your number of hinges in half at the roof level because you have one at top of column instead of two in beams to either side of column. Don't know if that mucks up energy dissipation at all, but again P&P doesn't seem too concerned. Think we'll look at just keeping column sizes and concrete grades high enough to keep our factored axial load under Agf'c/10 so we've got the codified way out but having P&P backing it up gives a bit more confidence that it's the correct way to go.

Thanks for the tip.