End plate connection

[JStructsteel]

Im following AISC design guide 39 end plate design, and for those that have it, Im following example 5.2-1

Im trying to calc my column web shear rupture strength.

Now, for the beam shear rupture strength, it calcs out teh beam flange strength and you check it against your shear. Makes sense.

For the Column web shear rupture strength, they seem to be checking the strength required against the value used to check the beam flange to the beam end plate. That ends up being the max calculated tension at the end plate, or 60% of the beam flange tensile strength. I think I get using that for the end plate and weld to end plate, but once that load is in the column, cant I calculate my shear rupture against the actual tension from the end plate, not the beam to end plate max values? Are they just doing that to be conservative and have the column be as strong as the beam flange/end plate strength. i.e. not have the column be the weak link?

I hope I explained that right.

 

[JoshPlumSE]

Look at page 46 (section 3.7.5) of Design Guide 39. I THINK this is what you are asking about.

The weld of the beam web to end plate in the tension region (taken as the portion of the web from the inside face of the flange to 6in beyond the bolt row farthest from the beam flange) must develop the full yield strength of the beam web in tension.

 

[JStructsteel]

Thanks. So based on that, then the column web should also be designed for the beam to end plate capacity, and not just the actual Pu in column from the tension load in the bolts?

 

[JoshPlumSE]

I believe the idea is that when the flange is at it's maximum tension there is a danger that the web weld could fracture since it's strain is basically the same as the flange at this intersection. Therefore, you need to make sure the web weld in this vicinity will survive until after the web itself begins to yield.

 

[JStructsteel]

Thanks. On a lightly loaded frame, this is forcing a much bigger column for one limit state. I would think that my column shear rupture would be the web area and the front flange area. Its using the web area only. wouldnt my shear rupture be something like this:

 

[JoshPlumSE]

I believe I misunderstood the original question.... What I cited shouldn't be affecting the column weld at all. I was talking about the beam web to end plate weld where weld fracture can be prevented by a slightly larger weld.

I hadn't actually looked up example 5.2-1. Now, that I've looked at it, I see the odd thing you're talking about.

The issue is where they say the following, right?

Check Cap Plate to Column Web Weld
The cap plate to column web weld required strength is conservatively taken as that for the column flange to cap plate weld..

If so, I think they're saying.... "What's the load demand on this cap plate? I don't know, let's just conservatively say this value...."

I'd argue that you'd want this weld to be strong enough not to fracture at this point (like the beam requirement that I cited), but how does that cause you to size up the column? If the column web would yield before the web weld fractures, that's good. So, I don't see why you'd need to size up the column.

 

[JStructsteel]

Ok, thanks. I was hoping I could use some judgement. My welds are appropriate, but the column shear rupture check is failing (if appropriate at all). Im still 5x my required tension, its that min check that is failing my column. I am comfortable ignoring it and design for my load.

Thanks for confirming my thoughts.