HSS welded OMF

[Prestressed Guy]

This is a question relating to the design of connections in a moment frame in a residence built under an SDS D seismic requirement and is required to meet the requirements of and Ordinary Moment Frame (OMF).

I have recently been seeing several of the engineers in my area that are designing steel moment frames for residential construction built entirely of welded HSS sections. The typical frame consists of two story HSS 6x4x3/8 tall continuous columns. The beams are also HSS sections 4” wide. The top beam is miter cut to the column and welded with a “complete penetration weld”. The lower level beam is welded to the face of the column with an “all-around complete penetration weld”. The members are sized to resist the demand load for the frame and it is assumed that this connection will develop the full capacity of the weaker of the two members. There is no analysis of deflection and the load is based on R = 6.5 because the building system is wood shear wall.

In my reading of the code this does not conform to of the FR examples in section 12-21 “RF CONNECTIONS WITH HSS” and certainly does not meet the limit state checks of an OMF.

Can you give me your opinion of the use of this connection?

 

[StructGuy123]

Why are the loads derived from an R of a wood shear wall system when a steel OMF is resisting the lateral loads?

 

[sandman21]

The R should be the lowest in that line so OMF. Drift needs to be checked many of these frames are assumed pinned at the base leading to large drifts. You can not truly get a CJP with HSS the same size the radius prevents this, must of the time it does not matter. Our office does not provide a miter for building moment frames. At the connection of the beam to column we provide a stiffener plate at the top and bottom flange.

 

[Prestressed Guy]

StructGuy123, I agree R should be 3.5 if it meets the requirements of OMF. If it is designed to a FR but does not meet the standards of OMF it should be designed with R=3 maximum.

Sandman21. I have not seen any of these calculations with a drift check. The base connection is pinned so they will have large drift. The welds do not have backing and both members are the same width so the CJP cannot be made. There are no stiffeners in any of the joints. As I see it, if these frames get anywhere close to column capacity at either beam, the unsupported webs will buckle and the frame will fail.

I was just stating what I see being submitted for construction by other engineers in my area. This type of frame is very common in my area and is routinely accepted by the building departments “because it was stamped by an Engineer”. I don’t agree with any part of it, I am just looking for the opinion of other engineers. I have started talking with several of the local building officials about this issue and would like the opinion of other engineers. The details are so similar between the different engineers that I suspect that one came up with the idea and the others followed suite.