Industrial building design 2

[RattlinBog]

Let me know if I should move this to the student section. I work full time at an industrial facility as an owner-structural engineer, but I'm also working on my M.S.C.E. part time. This post is technically about my plan B grad project, but I feel like it's relevant to any industrial structure design, too.

For my grad project, I'm working with my professor to take one of the large structural steel buildings at the place I work and design it from scratch with modern codes as a thorough exercise. Original structure was built in the 70s. My goal is to use the same overall building geometry but analyze and design it for ASCE 7-16, AISC 360-16, etc., and try to cover as many structural systems and components as I can--diaphragm, LFRS, purlins, girts, columns, floors, connections, foundations, etc. I'm trying to do most of the design by hand and only rely on software to double-check things. I don't have a lot of overall building design experience in practice (especially now that I'm in an owner role), so this project has been helpful to understand how individual components act together in a system, especially for lateral loads.

Would it be okay if I used this thread to ask occasional questions as they come up?

If so, here are some questions about diaphragms, purlins, girts, and eave struts. (I've read through and bookmarked several past eng-tips posts about this, but there are still a few questions in the back of my mind.)

Building info:
Ground snow load = 70 psf; basic wind speed = 107 mph
150 ft (north-south) x 456 ft (east-west) x 91 ft mean roof height.
Roof: purlins on trusses @ 24 ft O.C. (north-south)
Diaphragm: horizontal brace system at bottom chord of roof trusses (assuming roof decking attached to purlins is not designed as a diaphragm)
Lateral force resisting system (LFRS): 3 lines of braced frames (east-west); and built-up/braced cantilevered double column tower @ 24 ft O.C. (north-south)
Building is similar to a PEMB in that it's a massive structural steel frame with a lot of open space and no intermediate diaphragms, but all structural members are hot-rolled standard AISC shapes (no Z-girts or rod bracing, etc.)


Questions:
See attached calcs for purlin and girt design.

1. My understanding is cladding/roof deck will provide continuous lateral bracing for the exterior flange of girts and purlins for pressure wind loads. For suction/uplift wind loads, sag rods will provide discrete lateral bracing to the interior flange by preventing twist through a resistant force couple with the cladding/roof deck. Is that what others have seen? (I saw some eng-tips members from Australia and Europe mention bridging instead of sag rods, but I haven't come across that in the buildings I've seen locally.)

2. Is cladding dead load typically picked up by the eave strut through sag rods in tension? My understanding is that girts (assuming channels) are not designed for gravity loads in weak-axis bending and are only meant to resist wind loads in strong-axis. The gravity loads (cladding) are brought up from the girts to the sag rods to the eave strut.

3. Is metal roof deck typically ignored for diaphragm design in large industrial buildings? That's been my assumption so far, and I'm only relying on the horizontal plane bracing in the roof to bring lateral loads to my LFRS.

4. Are eave struts typically designed for both gravity (cladding dead load) and lateral (as a collector for bringing wind loads to LFRS) demands? Or is it better to have two separate members? Seems like two members might be warranted as the girts and LFRS bracing would be offset.

Thanks!

 

[RattlinBog]

Sorry, attachment didn't like "&" in the title. Trying again

 

[dik]

Is this a PEMB? If so, you will likely have a little difficulty in designing it from scratch... The designs are extremely tight and very difficult to replicate unless you have dedicated software.

-----*****-----
So strange to see the singularity approaching while the entire planet is rapidly turning into a hellscape. -John Coates

-Dik

 

[RattlinBog]

Not a PEMB. Here's a typical section

 

[dik]

Not even close to PEMB, thanks... You may want to look at your cross-bracing. I don't know why it is interrupted, and tension (and compression 'X') bracing in the lower levels may be required. With framing like that, it's my experience there will be a slight reduction in size due to increases in moment capacity. Are you using ASD, LRFD or LSD for design. I like using plastic design, but the structure doesn't appear to lend itself to that.

-----*****-----
So strange to see the singularity approaching while the entire planet is rapidly turning into a hellscape. -John Coates

-Dik

 

[human909]

RattlinBog, you get a pink star for such a polite and thorough post. I wish all people posting provided such sensible and well described set of questions.

I'll post my comments. I'm very experienced in these types of buildings both old and new. However construction methods do vary by locality.

1. My understanding is cladding/roof deck will provide continuous lateral bracing for the exterior flange of girts and purlins for pressure wind loads. For suction/uplift wind loads, sag rods will provide discrete lateral bracing to the interior flange by preventing twist through a resistant force couple with the cladding/roof deck. Is that what others have seen? (I saw some eng-tips members from Australia and Europe mention bridging instead of sag rods, but I haven't come across that in the buildings I've seen locally.)

I'm from Australia and that is EXACTLY how rods/bridging is considered. In Australia we used to use rods but that is largely surpassed by clip bridging to to the speed of installation. I would point out that our roof decks are normally only 0.42mm or 0.48mm thick. Which seems to be 30%-50% of the thickness used in North America. So designs are certainly different here. Wall here is ~0.35mm thick

2. Is cladding dead load typically picked up by the eave strut through sag rods in tension? My understanding is that girts (assuming channels) are not designed for gravity loads in weak-axis bending and are only meant to resist wind loads in strong-axis. The gravity loads (cladding) are brought up from the girts to the sag rods to the eave strut.

I normally ignore the precise load path of the cladding weight. I apply it to the columns. This is because:
-sag rods are not always present or under tension (at lease in the girt/cladding systems I use/encounter)
-the cladding is a stiff diaphragm at least relative to the cladding and girt weight. Thus it will maintain it's shape and transfer the loads to the stiffest points which as the columns (through the girts immediately adjacent to the columns.)
-'sag rods' or bridging as it is know in my locality is generally more providing twist restraint to prevent LTB

3. Is metal roof deck typically ignored for diaphragm design in large industrial buildings?

-I would absolutely ignore in my locality.
-A metal roof is often not stiffly attached to major structural elements at least in my locality. I'd consider it to be applying restraint to local elements roof purlins or trusses. But I wouldn't rely on it to transfer global loads.

4. Are eave struts typically designed for both gravity (cladding dead load) and lateral (as a collector for bringing wind loads to LFRS) demands?

Depends on the exact design but normally my eave structs my experience don't carry and gravity load apart from self weight. The provide a rigid tie for the top of the column both to ensure that it is not a cantilever column and to provide a path with wind loads.

 

[RattlinBog]

human909- Thank you! I try to be detailed but usually end up long-winded. Thanks for answering my questions. That helps a lot. For cladding weight, I figured the load would eventually end up in the columns, but I wasn't sure if it was standard to assume it went up to the eave strut first. For your typical eave struts, do you usually just have one member? Or do you ever have one cold-formed strut at the corner, and then a separate hot-rolled beefier member acting as the collector? I saw this figure below in "Elastic Design of Single-Span Steel Portal Frame Buildings to Eurocode 3":

dik- the bracing in the middle tower LFRS is interrupted for walkways, pipes, and other utilities through the 456 ft length of the building. All the horizontal struts/beams attached to the double columns have moment connections. I believe they added bracing to increase stiffness, too, so it's a hybrid moment/braced frame tower with some bracing interruptions. If I was doing this design from scratch (which I am, but only as an exercise) I would probably use as much X-bracing as I can, like you mentioned. The north-south LFRS definitely drifts more than the east-west braced frame LFRS.

For my grad project design, I'm primarily using LRFD but occasionally dip into ASD (strength, not stress) when I'm feeling lazy and checking service level deflections (all based on ASCE 7-16 load combos).

With framing like that, it's my experience there will be a slight reduction in size due to increases in moment capacity.

What do you mean by that?


Some additional info on the LFRS and a few extra questions:

Below is a plan view of the truss bottom chord bracing, which I believe is intended to be the only diaphragm in the whole building. Does that seem right? I see no other diaphragm if I ignore the roof decking.

Below is a typical east-west braced frame LFRS (3 total) elevation. Nothing too unordinary here. They provided struts and sway bracing at 5 locations at the top to brace the roof trusses and prevent twist. One thing that maybe could be improved is having sway bracing in every bay (or every other) so there's no doubt that each truss is braced/kept from rolling. Might be overkill, though. No questions on this one, unless someone sees something I'm missing.

Below is the north-south LFRS (20 total) elevation. I've highlighted what I believe to be the only structure providing real lateral stiffness in that direction. I can attach my work later, but in my first pass of hand calcs, I tried to analyze the north-south structure as a whole using the portal method. However, after estimating some moment of inertia for the trusses, center double-column tower, and the two outer columns, it was pretty easy to tell that the middle tower was taking 99% of the story shear. My question--does anyone have a different perspective on the north-south LFRS? Am I missing something?

P.S. I'm assuming the three floor levels on the right side of the double-column tower are not acting as diaphragms or are part of the LFRS...maybe that's wrong. I do assume they laterally brace the columns.

 

[human909]

human909- Thank you! I try to be detailed but usually end up long-winded. Thanks for answering my questions. That helps a lot. For cladding weight, I figured the load would eventually end up in the columns, but I wasn't sure if it was standard to assume it went up to the eave strut first. For your typical eave struts, do you usually just have one member?

Yes just one member. Pretty much as shown.

Or do you ever have one cold-formed strut at the corner,

I have a cold formed member at the end to provide support for the wall and the roof cladding as well as the guttering. I don't really considerer this a "structural" member, though it does need to have sufficient capacity to take the local wind loads which are usually higher at the building edges.

and then a separate hot-rolled beefier member acting as the collector? I saw this figure below in "Elastic Design of Single-Span Steel Portal Frame Buildings to Eurocode 3":

Yes I have a separate hot rolled member. Though I personally wouldn't use the terminology of collector, it doesn't seem to be collecting any load from the diaphragm. (Though in my locality we don't use the terminology of 'collector' anywhere.

 

[HTURKAK]

Yes..

- will you please post the end walls? Any vertical bracing at end walls ?
- Are the beams of the three floor levels on the right side rigid connected to columns ?
- Not sure if the middle tower was taking 99% of the story shear. Since the bracing at this stanchion column is not continuous and the column at axis B resisting with strong axis.. I would not underestimate the resistance of frames at three floor levels on the right side .

Just saying..






Use it up, wear it out;
Make it do, or do without.

NEW ENGLAND MAXIM

 

[RattlinBog]

HTURKAK- Attached are some original drawings of plan views, elevations (including end walls), and a few fabrication drawings showing the center double-column tower (stanchion column) and a typical girder at El. 1720'-0"

The end walls don't appear to have any additional bracing or lateral load resisting members. There are some wind posts and lots of girts and sag rods. I was hoping to see something else lateral-resisting besides the middle tower, but nothing jumps out at me.

The floor beams at the three levels are unfortunately just simply supported between columns. I have the erection/fabrication drawings (and ability to see in person) to verify. One shop drawing attached (MK 286D) shows a typical W36x160 that is between column lines E and F at El. 1720'-0". I can definitely see how the 3 floors could take some lateral load, but I'm not sure if that's how they were originally designed. The real structure obviously doesn't care what the design intent was, but it's hard for me to tell what's reasonable.

I believe the intent of the center double-column is to act like some sort of hybrid vierendeel truss with intermittent braces. I've never seen something like it, but that's what the shop drawings seem to indicate...the struts between columns are rigidly connected (flanges & webs welded). Do you see something different? I also have some original 1970s design calculations I can dig up. They're a little bit hit or miss as far as comprehensiveness.

 

[dik]

Nearly always use it... but we're not in a seismic area, and I'm not sure of what the consequences are.

-----*****-----
So strange to see the singularity approaching while the entire planet is rapidly turning into a hellscape. -John Coates

-Dik

 

[bones206]

I believe the intent of the center double-column is to act like some sort of hybrid vierendeel truss with intermittent braces. I've never seen something like it, but that's what the shop drawings seem to indicate...the struts between columns are rigidly connected (flanges & webs welded). Do you see something different?

Just took a quick glance at the drawings. It looks like the exterior columns were detailed with fixed base connections.

 

[RattlinBog]

bones206- See attached. Column line B (south wall) is a fixed base (type "B" base plate), but column line F (north wall) is a pinned base (type "A" base plate). Column lines A, D, E, and F are all pinned bases. Col lines B3 and C are type "C", detailed to take a lot of axial/uplift load from the force couple.

 

[hokie66]

RattlinBog,

I echo the comments of human909 about the thoroughness and politeness of your posts. You have found the right forum. Now you just have to decide whose advice to adopt, as there are always different opinions.

One thing which needs to be clarified. Metal roof deck and metal roofing are different things. Except in very small structures, which yours is not, roofing should in no case be relied upon as structural diaphragm. It is for shedding water, and that is a critical job in itself.

 

[RattlinBog]

hokie66- Thanks! I try to keep many friends and few enemies. I've gotten by in life so far.

Agreed on the roof. We have 1.5B roof deck on this building (roofing membrane over the top). It's screwed into the purlins every few flutes. Not sure if it was designed as a diaphragm or not in the 1970s. Photo attached (I would embed more images in my posts, but my internet security at work blocks that function...I also can't see anybody else's post-embedded images. I have to wait until I get home to see all the cool pictures)

Edit: hokie--I have a lot of old eng-tips threads bookmarked about girts/sag rods/bridging/bracing with your replies/inputs. They were helpful to read through! I wish I could see more examples of purlin and girt bridging (vs. sag rods) in the US.

 

[hokie66]

See page 10 of the attached manual for bridging as it is done in Australia. All part of a system. Each purlin manufacturer here has a similar system.

Yes, your roof deck and membrane above is typical in the US, but not in Australia. Metal roofing in Australia is not rigidly attached to the purlins. For concealed fastener systems, brackets are used which allow the roofing to move. For exposed fastener systems, the screws go through the ribs, not the valleys.

http://www.cadtech.com.au/Products/LYSAGHT/Zeds_Cees_Manual.pdf

 

[RattlinBog]

Looks like your attachment is missing.

I messed around with my Risa-3D model a bit and got things to work. I removed the three floor levels for now to see how the north-south (Z-direction) LFRS behaved by itself. See attached. (My model layout/geometry doesn't 100% match the real structure, but it's fairly close.)

I'm getting 2.2" of north-south drift(0.2% drift ratio) and 0.37" of east-west drift (0.03% drift ratio).

Model includes dead, snow, wind uplift, x-wind and z-wind (including eccentricity), and notional loads (per AISC direct analysis method). The total north-south (z-wind) at the diaphragm is 709k and total east-west (x-wind) at diaphragm is 188k.

 

[hokie66]

Sorry, now it should work.

 

[RattlinBog]

Thank you. That bridging is pretty cool. So it attaches to a clip at the bottom (to a grade beam/slab)? I'm guessing it's good for tension and compression?

My last attachment didn't work either. Here's the image:

 

[dik]

Thanks, hokie...

-----*****-----
So strange to see the singularity approaching while the entire planet is rapidly turning into a hellscape. -John Coates

-Dik