Flat Slab Edge column connections

To the first I would say that if you rely on frame action to get stability, you normally would accept some degree of believable fixity to make the overall lateral resisting system more slim. What can be a believable degree of fixity there, there are many opinions, but a nonlinear analysis revealing the degree of the cracked zone can give an idea of how you need to model stiffness, or you can check in different articles etc what is said about. Some go to design based in an entirely elastic scheme (even if with some reduced stiffnesses) with almost unnoticeable effect in serviceability and maybe something more at limit strength. Since for use it is feasible, it is done (may or not meet some particular codes).

When you rely in an entirely separate lateral resisting system, say shearwalls or braced frames, I find more acceptable to model as hinges the supports at exterior columns, specially in not highrise buildings where the lateral displacement may be reduced with such systems and so any effects of the notional hinge small. In spite of that here in Spain has been common (for framed, not particularly flat plate, buildings) to reduce much the moment at exterior supports respect that produced by elastic fixity. Again the visual effect in common dwelling buildings is nil, and you see there more "crack" due to the separate concreting of column and floor than mechanical cracking from overstress.

Respect transfer beams with important loads, yes, it is common to make use of hinges or reduced fixity at ends to ease some constructive details. You may have also cases with elastic fixity and well, as I have practiced and seen them have not resulted as bad as one could at first sight think. Taking unto account continuity also helps the lateral resisting system, for here, the buildings not being very tall, definitions of lateral resisting systems as separate entities are not as differentiated as in the US, and everything works together. I also have designed hinges for ends of steel multispan transfer beams to deliver forces to concrete through an anchor plate backed by an array of studs in columns, and bearing in sockets in walls, and again, no problem. The big stiffness of the transfer beams has much to say on this since it forces not much the respective housings other than in the devised manner. It is in the same manner that a shearwall protects weaker beams with ends designed as hinges, by diminishing the displacement excursions.

Yor last question is maybe where we may produce a more tolerant design, since we have plenty of support. You may design the node as simply supported, you, from your questions, are aware of the cracking issues, so it is a matter of preference and, for dwellings at least, minimum geometrical rebar should control the issue to limited proportion. In any case since the trend is to mandate quite significant minimum rebar anywhere and specially where tension appears, I have sometimes (much times) used elastic fixity atop RC walls, and I remember once it was to the surprise of some insurance reviewing party that was criticizing bottom rebar of the end spans without thinking I have done such thing.

Maybe after this account of my experience, the bottom line is that you can take almost the whole gamut of fixities and still have satisfactory buildings; that some may sastisfy some particular code or preference is an altogether different thing.

For me , i follow the same concept as you mentioned

Flat slab always re-distribute the stresses

in some cases where flat slab deflection is critical , i assumed fixed connection with column/wall and should design the column for this moment and reinforce the slab as well.

For RCC beam framing to column ,if no negative steel is provided beam would crack at that pint and hinge/pinned connection would form,this should be considered as hinge in lateral force resisting system as well

1. Unless the corner columns are very stiff they should not attract a great deal of negative moment to the slab. The exterior columns do need to be designed for this moment transfer.

2. If you are going to redistribute from negative moment to positive moment you will need to thicken the slab to account for the extra deflections.

At times, when I have had post-tensioned transfer beams I have extended the prestressing to the top of the slab as opposed to the centroid of the beam. I have used this to counteract the negative moment due to fixity. Have others used a similar approach?

I have also seen in the post-tensioning manual that a column can be assumed to be cracked and beam designed for positive moment. However, they recommend using closely spaced ties at the top of the column. I will try to scan and post that page from the PTI Manual.

For flexure, you can assume the ends pinned. But for punching shear at the columns, you have to design the slab to cope with the moment.

hokie66, I'd add to your sentence that for column design you need to include fixity as well, right?


For the original question:
If you "let" the slab crack with little or no top steel, I'd be worried about shear transfer - a vertical tension crack due to top moment bending sure seems to me to reduce shear capacity significantly.

JAE
By letting the column crack at the location, you are reducing the joint stiffness and not affecting the slab shear transfer. I am away from the office and will try to scan the page from the PTI manual that makes specific reference to a situation such as this.

See attached

slickdeals - yep - agree.

I was responding to the comments above that talked about minimizing (or leaving out!) negative SLAB reinforcement.
Suggesting a crack in the slab at the column, not in the column itself.

That may be a way of analyzing the joint, but I don't think slab-column joints crack in the column due to flexure. Column cracks do form, particularly in the top level, due to shrinkage of the slab, but not due to rotation.

Thanks guys

For a general flat slab, when doing edge column design, would you just allow for nominal eccentricity of 20mm as per BS8110 and use this moment? This is prior to doing a full scale FEM analysis and finding out what the actual moment transferred to the column is. I am only doing preliminary design.

If say I have 2000kN on edge and internal columns, can I design the columns for the axial load plus 0.02x2000 = 40kNm?

When checking punching shear at edge columns I do increase the shear applied by factors as recommended in the BS.

I’m a bit late to this discussion, but I agree with Hokie and JAE on all points.

For you latest question as to preliminary design, you need to be very careful here, If you size you columns for axial load only and you’re working on a talk building, you column will shrinkage and creep further than your core/shear walls. You want to design your columns and shear walls for strength but then also ensure the average stress is the similar.

I also recommend reading :
Manual for the design of concrete building structures to Eurocode 2
faq507-1574 (I am in the process of updating the FAQ's at the moment so take a look in a few days and the info should be better)


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