Beam to Column Joints in OMF 2

[EBF]

According to 2002 AISC Seismic Provisions, Sec. 11.2(1) FR moment connections need to be designed for the flexural strength of the beam defined as 1.1RyMp. I am using a WUF-W connection (FEMA 350 Sec 3.5.2), which has the beam flanges welded directly to the column flanges. For the CJP groove weld at the top and bottom flange, the tension capacity is 0.9FyA. How can this requirement be satisfied when the capacity of the weld is equal to the flange tension capacity but the required strength is the beam capacity increased by 1.1Ry? For a W18x 50 beam, 1.1RyMp = 1.1*1.1*50ksi*101in3 = 6110 in-k which results in a flange force of 6110/18in = 340k. The capacity of a CJP groove weld would be 0.9FyA = 0.9*50ksi*7.5in*0.57in = 192k. Please explain where I am going wrong.

 

[JAE]

I don't think you are going wrong - just that you are finding that using a light, deep member creates difficulty in ensuring that a plastic hinge will occur away from the connection.

Do the same calculation with a heavier W18 - say a W18x175
1.1RyMp = 6413 in-k
Flange force = 6413 / (20.04-1.59) = 347.6 kips

Weld - .9 x 50 x 11.375 x 1.59 = 813 kips

So what you have is a too-deep beam for your situation. Otherwise you will have to build-out the end connection similar to the examples shown in the AISC Seismic Commentary (See Figure C-11.1 in the 1997 yellow book) where a haunch, stiffening rib or cover plate is used.

 

[EBF]

I'm not getting the same values. For a W18x175 (Fy=50ksi and Z=398in^3):
1.1RyMp = 1.1RyFyZ = 1.1*1.1*50*368 = 24079in-k (not 6413)
Flange Force = 1305k > Weld capacity = 813k
Is that right?

 

[JAE]

Ooops - I was using Zy

1.1RyMp = 1305 flange force - you are right - now I'm flumoxed.

 

[JAE]

OK - here's my take on it:

I believe that post-Northridge design requirements try to force the plastic hinge away from the column flange by forcing you to reinforce a short portion of the beam at the end (per my last paragraph above). This then would allow you to satisfy the 1.1RyMp = Mu limit.

But the 1.1RyMp limit also includes the statement "or the maximum moment that can be delivered by the system, whichever is less." Many times you floor or roof diaphragm cannot transfer enough lateral shear to develop the beam to 1.1RyMp so you can use the smaller value if this is the case.

Because of this insistence that the hinge occur in the beam and not the connection, some proprietary connections have been developed (SidePlate connection and the RBS -reduced beam section or dog-bone) to reduce the beam's 1.1RyMp and then not require a built-up connection, which is more expensive.

 

[EBF]

So what it comes down to is that unreinforced flange connections (such as the WUF-B, WUF-W, and Free flange connections) can not be used even though they are pre-qualified by FEMA?

 

[Taro]

EBF,

Note that the FEMA 350 OMF is the same as the AISC 2002 Seismic Provisions IMF. Only the name has changed. So just call the system an Intermediate Moment Frame and avoid the whole issue.

 

[JAE]

Taro - I was hoping you'd weigh in for the rescue -

EBF - you can use those connections IF the maximum force that can be dragged into the system is less than 1.1RyMp (under the OMF - without considering Taro's comment).

 

[EBF]

Thanks for all the input, its been a big help. From this discussion I have gathered that these detail requirements in the 2002 Seismic provisions for OMF (as well as the new limits on height mentioned in the commentary section C11.1) are put in to force the use of a more ductile system, i.e. SMF or IMF. Do you agree?

If I understand correctly, according to the 2002 AISC Seismic Provisions, the OMF connections don't have to be prequalified, they just have to meet the 1.1RyMp strength requirements. And the IMF is basically an OMF per FEMA 350 with a prequalified connection, right?

 

[seismiczone]

Son't forget that AISC 2002 may not be adopted in your jurisdiction. You may be stuck with 1997 code.

 

[EBF]

That brings me to another issue. I work in the San Francisco Bay area, where the governing building code is the 2001 California Building Code. This is based on the 1997 UBC which refers to the 1992 AISC Seismic Provisions. In my few years of experience, I've found it to be common practice to use the 1997 Seismic provisions for steel detailing, even though it is not what is referenced in the CBC. Have other people had the same experience?

 

[seismiczone]

Which brings to mind another issue! I have received plan check comments from San Diego on a two story steel frame residence. Evidently the City has a memo that allows non-qualified connections if the R value equals 1.0 in calculating the shear for non-qualified connections. I guess the calculated force is increased 4.5 times at the connection as a safety factor guarranteeing the connection functions in the elastic range. No testing is required. Does this make sense?

 

[08zY]

hello, people im just new in this work as a structural engineer of my own company.still i considered my self as a junior structural since i have designed only up to four storey building and most of them are located not in a very critical areas that needs a serious computations.
Now my problem is everytime i start to compute still my basic problems is the assumption of loads. it all started when i used STAAD Pro. because the results is always higher than what my own computation. Is it my loads that im using is not enough and far enough to what STAAD standards have?

 

[JAE]

08zY - you should start a new thread and not add to this older thread. You may get more responses that way.