capacity check - shear links 2

[oneintheeye]

Forgive me here but this is relevent to a British standard clause and I realise most here are US.

BS8110 (Concrete) states that all tension bars in a beam must be within 150mm of a vertical leg. It also gives other min dims. I don't know if you have similar clauses in US.

My question is this;

I am looking at an existing stucture for some increased loading, the check calls for links for the shear. The area is fine, however the link arrangment does not meet the detailing clause above. I am guessing that maybe in an older code version this was acceptable or the original design didn't need links so they put in the minimum area (albeit not to BS8110 detailing rules, but they could in theory be omitted totally).

To my mind the links are required to halt any shear crack propogating down to the main bars and reducing the capacity of the section be breaking the bond between the main steel and the concrete, Leading to failure. Hence the spacing only 150mm from the tension bar. Therefore in my assessment the non confoming detail will mean the beam in idadequate in shear due to the detail even though the area is above that required.

Any views?

 

[ishvaaag]

Whilst in a walk I realize that what above I have called a "cosmetic" moment it is not so for the reduced number of longitudinal bars being accounted. The program is for shear field action and then needs vertical and horizontal reinforcement available to sustain the "strut" action hence a severe reduction on steel available for this action reflects inmediately on the shear available. That means that if you consume all longitudinal steel for moment, the model can't return much above plain concrete shear strength, so don't ring the bells for now. Soon after using a bit more the worksheet will post what found.

 

[ishvaaag]

For the moment disregard the printout given above for the moment was excessive for the allowed longitudinal steel.

 

[ishvaaag]

The worksheet is for optimization and showing scarcely sensitive to the change in longitudinal steel. Have other variants of the same that could be, but rather try another approach with others about shear to see some cooked solution within my tools.

 

[ishvaaag]

It would be useful to have Mu Vu concurrent for design. For the moment, accounting with the 9 16 mm diameter bars at one side, if crack control is to be respected, the reinforcement would be allowed to take a bit less than 120 metric tons in shear by the spanish EHE-98 (former) code ruling RC. If you take 5 bars out of these, this is reduced to around 85 tons (factored). This is for fck=30 MPa that would allow for fcd=20 MPa.

Estimates according to Rahal give even lower than by the previous evaluation.

One evaluation (on Gaetano@Puleri at ACI SJ) goes to 480 metric tons for fc=30 (without fi reduction). Maybe not as dissimilar as apparent, they surely are targeting the actual final value (put, as quoted above, by Nawy, about 20% of fc, here shown less percentage, depth effect likely) and so not comparable to code practice. As much difference we must expect between what we design and the limit strength, 3.5 for ordinary checks may be quite likely and over 4 for connections and risky checks, like fragility in shear. This is not a code check, but makes clear that one shouldn't expect THIS member fail quickly in shear.

A check as per ACI as I have it in a render of it would accept 190 tonne factored shear.

So you have a whole gamut of opinions about your member.

Hope this helps.

 

[InDepth]

Herewegothen. For the reinforcment layout, did you pull these from a drawing? Did you field verify if there were additional links installed (maybe a construction change that wasn't documented). Maybe some non-destructive testing is in order.

On another note, isn't there an exception in the code that allows no stirrups to be installed if Vu < 0.5*phi*Vc (See ACI 318 11.5.6.1). Also in ACI 318 sec 7.11, you should pay close attention to the wording which activates the tie spacing requirements for "compression reinforcement in beams." If you aren't using the longtudinal bars for compression, then you don't need to meet thetie spacing requirments.

If all else fails, propose strengthening (vertical epoxy dowels, fiber wrapping, ect...).

 

[ishvaaag]

Also, if a shear field was of order, the skin reinforcement would be counted.

 

[oneintheeye]

in depth, no we have not feild verified but I have a couple of drawings showing the detail, one from the client one from the original designer. Also the calculated shear stress requires designed links i.e. is above the lmit you show to leave out links. I believe I have a shear approching 1200kN from memory.

Field chack is the next step, the structure is only rented so any remedial work opens up a new can of worms

 

[ishvaaag]

Here go 3 worksheets, one that of Vijaya Rangan method used for optimization, assuming the 4 16 mm diameter bars are available for shear-field action. The two others follow Collins&Mitchell estimate for shear in one of their books. The first would allow for Vd=1270 kN, and the two others may rate the concrete and stirrups contribution at 980 kN and 1448 kN, hence making a shear load at 1200 kN an unlikely problem.

I recently downloaded the freely downloadable RESPONSE series of programs and may use your case to try its use.

 

[ishvaaag]

Collins 1

 

[ishvaaag]

Collins 2

 

[ishvaaag]

I installed my RESPONSE 2000 copy (a program by Evan C. Bentz at the University of Toronto, directed by Michael P. Collins) and ran your problem. You can download a series of 4 programs, Response 2000 one of them at

http://www.ecf.utoronto.ca/~bentz/download.htm

In the first page note your section (forfeit the loads, these are not those later applied). Note that a 15M bar is accepted to have 200 mm2 section at the program.

 

[ishvaaag]

Then see some output for your shear in the following printout.

 

[ishvaaag]

And then see how to 226 kN·m behaviour goes nicely.

In all, even when the program is for shear-field action and so not the best choice where strut and tie schemes are, note that by this MORE realistic assumption than a pure Mörsch scheme there is still no cracks at your section, and the implied deformation has not even started to pass stress to the stirrup. So this is another favorable opinion on that the section will behave well.

You see in this model I have counted the 9 bars atop and at bottom, and not included any skin steel. I plan to enter the 4 above 4 bottom plus some skin steel case later; if something negative develops I will let you know, otherwise give this free nice program a try, it takes 15 mins to get these results.

 

[oneintheeye]

thanks very much, Ill look into it

 

[oneintheeye]

Update. I have inspected the structure. There is no visible cracking of the effected beams. Of course I do not know the actual loads applied to the structure or if they are any where near the design loads but at least I haven't seen any visible cracking due to shear.
I am awaiting further loading information in order to proceed.

 

[JAE]

herewegothen,

A few thoughts here - I think what hokie66 suggested above might be the direction I'd take - look at a reduced Vc based on what you DO have in terms of longitudinal bars contributing to shear strength.

In the US (ACI 318) I don't believe there is any provision similar to yours where stirrups must be spaced out across tension bars. Compression bars yes. I've always added additional links(stirrups) for wider beams though...just for feel good reasons.

The shear provisions for slabs vs. beams is based upon the idea that slabs can transfer load across slab widths and thus add to some redundancy. Beams are single, non-redundant elements so more conservatism required.

One "fix" idea: Could you drill vertically through the beam and install vertical bars in grout/epoxy/adhesive?...or perhaps plates top and bottom? Might be a lot of work and you might have struggles with hitting the longitudinal bars but since you said you were "stuck" this fix came to mind.

 

[hokie66]

Did I say that?

 

[JAE]

hokie66 - I must have been smoking something - I think I took your statement:

"If it is a wide, flat beam, you should be able to use the one way slab provisions rather than the beam provisions"

as using slab provision vs. beam provisions and maybe getting it to work....I was probably half asleep too.

 

[cntw1953]

I think ACI has questioned the effectiveness of a beam wider than the column. Just from memory, check it out.

 

[hokie66]

Huh? More often than not, beams are wider than columns, seismic requirements not controlling. They shouldn't be the same width, as it makes the reinforcement clash.