Checks Required at Concrete Column Transfer Slabs (Walking Columns) 5

[KootK]

Please review the sketch below (and attached). I am curious what others do to check slabs at column transfer locations. I'm thinking of "walking" columns in thinner slabs but the discussion is equally germane to full blown podium style transfer slabs as well.

The options:

1) Individual column punching shear with small moment.
2) Combined column punching shear with large moment.
3) Shear friction in the slab between columns.
4) Bearing stresses.
5) P x e moment in slab.
6) Vertical seismic acceleration effects.
7) One way shear between columns, similar to #3 (forgot to draw this)

I believe that the following checks are fairly standard: 1,4,5,6. Combined punching shear (#2) is a KootK invention as far as I know. Shear friction (#3) is standard practice at one of North Americas top tier structural firms. I feel that it's unnecessary as I can't imagine that ever governing over the diagonal shear failure modes (punching & potentially one way).

So, what do others check? And I'm not looking for FEM solutions. And I know it's awful.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[KootK]

Shear stress needs to be checked around the entire area, after that it does devolve into a how. If there is a way to accurately model/calculate the unbalanced moments and axial loads then you could look at the d/2 stresses away. The big vs little section is a how argument isn't it, it's how do you get what the true stresses are (since it seems like we agree that they are what they are, #1 and #2 are two guesses at 'how' you find them - only one is accurate).

To me, this sounds like simply evaluating shear stresses at all locations and designing for them. And, whether the stresses come from FEM or some old school hand method, we're talking about one-way shear stresses. I believe that two way shear is fundamentally different. From a code perspective, we don't just check one way shear on a square shaped perimeter around our columns. We actually check for a different stress.

My gut tells me that the difference is about proportion. Two way shear makes sense for stocky things that will move as rigid bodies and do a lousy job of redistributing shear. One way shear is for bendier stuff that can redistribute shear stress a bit. That's just my theory though. I've never heard any explanation for the difference in stress other than "it tests out differently".

My big versus little argument is about where the transition ought to be between where a one-way shear stress check is appropriate and where a two-way check makes more sense. My concern with mode #2 is that there may be a section of slab encompassing both the upper and lower columns that might still be deserving of a two-way check.

I agree with one way shear entirely and that it's a little bit fuzzy how far away you have to be before you spread that out

I'm glad that we agree on this and am simultaneously terrified. This would govern thickness in the vast majority of cases.

It is not the same but similar to a question I asked recently about 1 way shear in a mat foundation. Several responses opined that it's impossible and only the full mat width needs to be checked but I disagree.

I remember that thread as I was the primary responder. Only one commenter suggested that only a full width shear failure was possible. And, at the risk of offending that commenter, I still find that notion implausible. Suppose I build a suspended slab the size of the USA spanning between the eastern and western sea boards. If I drop twenty foot long a N-S running shear wall over Nebraska, am I supposed to believe that they'll be feeling that as one way shear in Fargo and Houston? Fat chance. There's a limit.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[KootK]

The difficulty I have is how to account for the fact that the direction of the shearing stress switches somewhere along the perimeter. I see the hunk of slab as more of a beam trying to rotate about it's centroid then I do as a lump of concrete getting punched out.

Does this clear it up Jayrod? The diagram at the right. You could have a chunk of concrete loaded entirely in flexure, with no axial load, and still have a punching shear failure. In fact, the situation that we're discussing is probably getting close to that.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[KootK]

As for number 7. How do you determine the length of the one way shear failure plane?

Precisely. I've seen no guidance on this. I would think some multiple of the slab thickness would be a good rule. If only I knew what that multiple was... I know Bookowski would like to know too.

I think it is likely that it may start failing in one way shear directly between the columns, but once the cracking reaches the edges of the columns it would need to turn into two way shear. There is no way it is going to cause one-way shear for the whole length of your slab.

Agreed. In my original sketch, I mentioned that I thought that a shear failure initiated as one way might morph into an unzipping style punching shear failure around the perimeter of the column. Whether or not that's a possible failure mode is an important part of that discussion.


The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[jayrod12]

I would say it should be considered as a failure mode. But my guess is that by the time you check the individual punching shears those likely govern except in extreme cases as the one you are illustrating. This example is an extreme case and highly unlikely. If this amount of offset was in the original plans we all (at least I think we all) would find some way to use more concrete to transfer the load into the offset column below.

But that's just my thought.

As for the diagram you posted, that is what I envisioned however, your punching shear perimeter on the original problem is not only unbalanced moments, you have punching shear stresses in both up and down orientations from moment and column loads. So when I imagined the total shear stress diagram there would still be an inflection point where the total shear stress is practically zero.

 

[KootK]

This example is an extreme case and highly unlikely. If this amount of offset was in the original plans we all (at least I think we all) would find some way to use more concrete to transfer the load into the offset column below.

It's an extreme example in all forms of work other than residential towers. Many of those will have this condition for light loads up at the penthouse and for heavy loads at the transfer slab near grade. Having checked these conditions numerous times I can tell you that the one way shear check, as proposed, always governs until you get some meaningful separation between columns.

however, your punching shear perimeter on the original problem is not only unbalanced moments, you have punching shear stresses in both up and down orientations from moment and column loads. So when I imagined the total shear stress diagram there would still be an inflection point where the total shear stress is practically zero.

This sort of depends on how you see things. As I mentioned above, I've been assuming that the difference between a one way shear appropriate situation and a two way shear appropriate situation to be whether or not I expect that the shearing block will behave more or less rigidly. If it is rigid, then you can aggregate the effect of all the moments and axial loads into a single, representative moment and axial load (like we do with eccentric footings).

With a single moment and axial load on an assumed rigid body, I feel that you'll get diagrams like I've posted above. Whether or not the shear crosses zero and flips sign would depend on the ratio of moment to axial force. In these transfer situations, the moments tend to be rather high and the only net axial force is the contribution from normal slab support loads. As a result, I usually do see a punching shear stress diagram that crosses the zero line.

Thanks for your endurance.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[paddingtongreen]

This has been an interesting discussion. The load, P is nearer the lower column than the far end support so the lower column picks up most of the load. The upper column's fixity prevents rotation of the slab, moving the point of contraflexure further from the supporting column, thus moving some of the floor reaction from the far end to the supporting column. The slab between the columns is subject to moments at each column, leading to your #7 shear. I would be tempted, nah, I would definitely use, strips that carried the load, There is a limit on how much slab would be carrying it, the width of the column plus two or three thicknesses perhaps, not much more. If this can handle the moments and the #7 shear, well done!


Michael.
"Science adjusts its views based on what's observed. Faith is the denial of observation so that belief can be preserved." ~ Tim Minchin

 

[KootK]

With respect to the one-way shear failure (#7), I'm attracted to the notion of running stud rails between the two columns. Ties for thicker slabs.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[jayrod12]

On the topic of stud rails. You should check out the ones by peikko. Pre - engineered as far as I understood, easy to site modify for rebar conflicts. Otherwise the same as all other

 

[KootK]

Neat.

1) do we want the spacing to be field adjustable? That makes me a bit nervous.
2) how do you mean pre-engineered? You supply just forces to supplier?
3) how do we feel about top side install? At first, I thought it would just be another cause of top steel -- and KootK -- depression. Now, I wonder if it might actually be better from the top. From a mechanics standpoint you want the stud head at the top of the tension steel. And having it higher is really of no benefit.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[bookowski]

Definitely on the ties or rails for one way, this is how I have done it. Usually ties since these are typically at least a bit thicker slabs.

 

[KootK]

But are you laying out your studs and ties to reflect the
two way check (#1) or the one way (#7)? I've never heard of the latter. And it would take a TON of studs in many cases.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[bookowski]

One way definitely, just was doing this last week in fact on project similar to what you were describing as typical (upper floor setback with three levels above setback).

What I do is show a line with a light plot style and dashed over a width and actually tag it with a beam tag. I usually also provide a section because I know it's likely to get missed/confused, I figure that beam tag + section mark should make it obvious. Basically I build a beam in slab.

 

[KootK]

Beam in slab... yes, that's exactly it. So what's it gonna take to get a clip from your plan so that I can see the representation? When I imagine it, I worry that it clutters the presentation of the other rails that may be installed for more conventional reasons.

Also, did you us Vc I your stud design or just Vs alone?

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[CELinOttawa]

I must be missing something... I see the offset column as a source of vertical load and moment, as well as lateral in EQ case, into the slab. The design of the slab is for the shear, moment, etc loads imparted. The design "stip" of the slab taking this is the perpendicular width of the slab plus 2d each way (standard punching shear provision).

What have I missed? Why is this causing a wide-reaching debate with some of the smartest folks around here.

Note: It is late, I am tired, and I just would rather ask and feel stupid in the morning than not ask and potentially miss an opportunity to learn.

 

[KootK]

@CEL: my intention with the thread was twofold:

1) I wanted to take a survey of what folks are checking in this situation and, hopefully, get some new ideas. While the group here seems to be pretty in tune with one another, I know that there is considerable variation in practice. My impression is that most folks just do checks one and five. Obviously, I seem to be a lone wolf when it comes to check number two.

2) In this situation, it is not at all clear to me whether one way or two way shear checks are appropriate. I wanted to get some input on that. My gut says -- as you seem to -- that it's predominantly a one way situation. However, I still "feel" the need for check number two. And that strikes me a permutation of punching shear.

So, now that you're here, which of the seven proposed checks do you support? Any additional ones? Why do we use different stresses for one and two way shear? Just different test results? I feel that it has something to do with the rigidity of the section that punches through and the capacity for shear redistribution in plan.





The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[paddingtongreen]

This is the first time I came across stud rails so I had to look. For shear resistance, obviously. The rail in a formed slab would provide embedded plates.

With all of those studs, the rail must be part of the reinforcing? it would seem so unless it can be rationalized away. If it is visible it can't be relied upon because the MEP butchers will surely chop some of it away. However, it seems to me that it must be included in the 75% of balanced reinforcing because it most will be undamaged.

Michael.
"Science adjusts its views based on what's observed. Faith is the denial of observation so that belief can be preserved." ~ Tim Minchin

 

[KootK]

Analytically, the rails aren't considered part of the flexural reinforcing. No doubt it participates locally but, since the rails aren't continuous over columns, they aren't relied upon for strength.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[CELinOttawa]

Okay... Well to start I don't see a low and a high moment, but rather the one moment caused by the offset of the columns. Are you meaning they in one case you accept reduction by returning moment of the slab?

As.for the checks, I also don't see the combined column punching shear. This feels a bit like a doubling of forces - Personally I think you've muddled the source of a load with the design of how to resist said load. I stand ready to be corrected...

The check of the shear through the slab is the core requirement. All else is secondary, even the flexural case. I doubt flexure would govern many of these, and even if it did, it is unlikely to be more than theoretically governing.

I'd be a 1-5-6&7 guy... With a long term creep check thrown in as well if the propodtions of the spans made me feel it was warranted.

Lots of engineers don't think of creep often enough; I've seen some very odd situtions with creep, and have learnt enough respect to keep it in mind.

I also think I'd be likely to request a drop beam to I make this all easier to do...

 

[jayrod12]

1) do we want the spacing to be field adjustable?

In reality all spacings are field adjustable it's just a which welding certificate is required to modify them. The peikko ones don't require anything beyond a basic welder's cert (which means joe blow and his mig welder can modify it if required) as the rail part is just a place holder and not an structurally required part of the system.

2) how do you mean pre-engineered? You supply just forces to supplier?

That's how it was explained to me, but it could of just been the guy blowing smoke.

3) how do we feel about top side install?

I agree with your new understanding, I think it's better from the top.

 

[KootK]

Thanks for your 2 cents CEL.

Okay... Well to start I don't see a low and a high moment, but rather the one moment caused by the offset of the columns. Are you meaning they in one case you accept reduction by returning moment of the slab?

As other thread participants have rightly noted, there are several sources of moment at the joint: 1) the column axial load eccentricity 2) the column end moments 3) the slab moments. These all exist and, in my opinion, are very difficult to estimate accurately. For the sake of this discussion, I've been trying to focus on the P x e moment as I believe that dominates the situation.

Lots of engineers don't think of creep often enough; I've seen some very odd situtions with creep, and have learnt enough respect to keep it in mind.

Interesting. Can you elaborate with reference to this particular situation? What do you see taking place on the creep front? Progressive rotation of the joint and exacerbated punching shear?

I also think I'd be likely to request a drop beam to I make this all easier to do...

No doubt we'd all want this. However, construction economics and intra-firm competition steer things towards a flat plate solution.

I also don't see the combined column punching shear. This feels a bit like a doubling of forces - Personally I think you've muddled the source of a load with the design of how to resist said load. I stand ready to be corrected...

Yeah, with respect to check number two, either you're all right and I'm crazy... or the reverse. Probably the former. I'll take one last stab at trying to sell the combined punching shear check. Consider the sketch below which depicts two situations:

1) The offset columns where the moment is P x e and it is dubious whether or not a punching shear check is required. Mode number two above.

2) A stacked column situation where the columns and slabs form moment frames for lateral resistance. Moment is still P x e but, here the need for a predominantly rotational punching shear check is obvious.

How about that? Anything? Does this resonate with anyone else???

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.