Footing Uplift 5

[slickdeals]

Folks,
When you calculate the uplift load on a footing, the code requires the use 0.6D + W. Assume this gives a net uplift of 30 kips.

When you calculate your resistance to uplift in terms of the footing weight and the weight of a truncated soil pyramid (based on a 30 degree angle), do you use a 0.6 factor on the uplift resistance and compare it to the above calculated uplift of 30 kips?

It seems like double dipping (very conservative) to use a 0.6 factor on the resistance side also. Thoughts?

 

[slickdeals]

I agree, we calculate the wind loads for structural uplift using MWFRS. The wind pressures drop drastically when you are >2h measured from the leading edge.

 

[ash060]

I asked one of the code writers for ASCE 7 this question, about very larger uplifts on the footing and if that was the intent of the code he said it was not, and he gave some guidance on alternatives.

One thing is that for alot of these one story buildings typically the columns are steel. The anchors rods are designed for the uplift from the column, but not to engage the entire weight of the footing. So the largest the footing should be is the maximum the anchor rods can pick-up. Adding more concrete than that is just a waste because the column will pull-out of the footing before all the weight of the footing is engaged.

For example the net uplift is 25 kips and the weight of the footing and soil on it is 40 k. What is the point of that if the anchor rods are designed for 25 k, any more foundation than that is waste. The 40 k will never happen.

So I have started designing for the maximum load that the structure can impose on the foundation, and not some number that the foundation can never see.

If Dorothy's house starts flying by my window it is time to start make our foundations larger, but until than maybe some engineering judgement is all that is called for.

 

[slickdeals]

@ash:
And the hawk-eyed code officials in South Florida are cool with it?

 

[structuresguy]

@ash: I don't understand your logic. If your net uplift is calculated by the appropriate Code mandated load combinations (say 0.6D + W), then your anchor rods will need to have a HIGHER capacity than just the actual weight of the foundations.

For example, if the gross uplift is 50 kip, and the superstructure dead load is 10 kip, the reaction at the foundation would be R=(0.6*10 - 50) = -44 kip. So the foundation would need to weigh (0.6*Dfdn) = 44 kip => Dfdn = 73.3 kip. So the anchor rods would need to be able to transfer all of this 73 kip load, since you are relying on all of it being there to counterbalance the uplift and satisfy the Code mandated load combination. So the anchor rods should actually be designed for MORE than the weight of the foundation, not less.

 

[ron9876]

@ash: don't forget that the anchor bolt design has material safety factors included so they probably could pick up the full weight. Not saying that I agree with the requirements but I don't see how you could argue that you met either the intent or the letter of the code if you don't use 0.6D. I think if you were ever challenged you wouldn't be able to defend the position.

 

[ash060]

@structuresguy: Anchor rods are needed to keep the column down, not pick the footing up

@ron9876: I met the intent, but not the letter. The footing is not going to uplift. If it is a failure that has never happened does it apply?

@slick: They have never asked

I use 0.6D for all the superstructure. The main thing I am trying to point out is that sometimes the code should not be defaulted to over engineering judgement. Anyone can follow the code, but not everyone is an engineer.

 

[cvg]

you guys are amazing! you mean that you don't always have to meet the minimum requirements of the code because you are an engineer? Than who was the code written for, all those non-engineer building designers?

I am a little hesitant to travel to south florida now. I expect that big hurricane that has never happened yet, to hit about the time I step foot off the plane...

 

[rscassar]

8x8x4' seems a little excessice for an interior column in retail space, even for hurricane zones. What I do is tie the slab into the footing and provide top steel in the slab so it can cantilever from the footing. It is also a good idea to match the size of the anchors with the mass of the footing to ensure continuity of the system.

 

[a2mfk]

I have spent my whole career as a structural engineer in Florida, mostly with low rise structures where this whole footing uplift deal is a big issue in design. I am relatively conservative as a structural engineer, but I see the two sides of this coin.

1. CODE:
I have argued there should be a small note added into the code to allow the use, at the discretion of the engineer of record, to use up to 100% (ok, maybe 90%) of the weight of an engineered slab and/or foundation.

But right now, if we are talking about what "code" says, I think they all say 0.6D or similar. And we all know how perfectly written codes are (tongue firmly implanted in cheek). Its been my understanding that the 0.6 dead factor is because the majority of dead loads have unpredictable weights that may change throughout the lifespan of a structure. A foundation of course does not fit this description, and thus in my opinion, it does not fit the intent of the 0.6 reduction. I do not believe you would violate the spirit of the code in this sense, for whatever that is worth I am not sure

2. PRACTICE
I have been fortunate to perform forensic investigations after Hurricanes Charlie, Jeanne, Francis, Katrina and several tornadoes. The only footing failure from uplift I saw was arguably a scour/storm surge failure, but a front porch post embeded in a small cube of concrete was probably about 50 feet from where it was built (on the bay in Biloxi). I have seen lots of roof failures, and structures ripped from the slab and foundation. Even a partial roof cladding failure will drastically reduce the uplift load getting down to a footing.

I think we are all not being realistic if we think that wind pressure in a one story building will pull a footing out of the ground by punching a whole in the slab. So the building envelope would be perfectly intact- roof, windows, doors, etc. which would allow the wind pressure to continue to keep increasing? And no element along the load path would fail prior to the foundation being literally pulled out of the ground? If this happens, we have bigger issues to deal with, like the survival of the species.

I guess that would be my argument for the code change to allow 0.9D for foundations and slabs....

 

[BAretired]

If you change 0.6D to 0.9D for ASD, you would have to change 0.9D to 1.35D for LRFD. I don't know how that can be justified.

BA

 

[ron9876]

Andrew was in excess of the design loads that were in place at the time (and now). I was with a firm that was involved with the review of the damage and there were no discussion of footings that failed in uplift. There are many older buildings down here that didn't consider uplift on footings. 3'x3'x1' footings didn't fail in uplift.

@ash the lawyers would eat you up. You wouldn't have a defense. We don't have to agree with the code but when we put our seal on a set of drawings it means that to the best of our knowedge the design complies with the code. We all know that the number of handicap parking spaces required at the mall are excessive because we have never seen them all filled ever. I don't see where there is any difference here.

 

[slickdeals]

We all know that the number of handicap parking spaces required at the mall are excessive because we have never seen them all filled ever. I don't see where there is any difference here.

I see a big difference because that is probably not a life safety issue.

However, I agree with you that if you certify a set of drawings as meeting the requirements of a building code, then you cannot choose to exercise engineering judgment. Probably there is no defense for that.

On the contrary, every building that is built is not built in 100% accordance with the code.

 

[BAretired]

The following comment was made earlier:

The piles have a safety factor from the geotechnical side. Then you add the 1/0.6 safety factor for the pile loads and you have safety factor x safety factor. There should be some type of allowance for this condition.

I don't think there is any doubling of safety factors here. A friction pile will have a skin friction value in pullout which is determined by the geotechnical engineer. Total resistance to pullout would be 0.6*Dp + Skin Friction where Dp is the dead weight of the pile. The skin friction value is not modified by the 0.6 factor.

BA

 

[ron9876]

@slickdeals yes but they had better be closer than a 67% overstress. There would be no defending that position.

BAretired my understanding is that the geotechs calculate the ultimate capacity of a pile and use a safety factor of 2.0 to determine the allowable capacity. That's what I meant. So you have the 2.0 safety factor and the 1.67 safety factor.

 

[BAretired]

ron9876,

You have the 2.0 safety factor on pile friction and 1.67 safety factor on pile dead load, but you do not multiply the two safety factors together. What allowance would you make for this condition?

BA

 

[ash060]

Does anyone agree that the code is excessive in this case besides me?

 

[BAretired]

I do not agree. The code seems reasonable.

BA

 

[ron9876]

@ash: absolutely.

BAretired: you reduce the dead load of the building by 0.6 whci is same as a safety factor of 1.67 and then you resist it with a pile with a safety factor of 2.0.

 

[abusementpark]

I agree with ron9876 here. There are many instances where there is a doubling of safety factors.

 

[BAretired]

ron9876,

You should have been a politician!

Say the total uplift force is F[↑] and the total dead load acting on the pile, including its own weight is D.

Say that skin friction failure load of the pile is Fs, so the geotechnical engineer gives you an allowable value of Fs/2.

Now, F[↑] - 0.6D = Fs/2

or F[↑] = 0.5Fs + 0.6D

There is a safety factor of 2.0 for the skin friction and 1.67 for the dead load. It is reasonable that a higher safety factor should be used for skin friction because there is a higher probability of going wrong than simply calculating the dead load. But we are not multiplying the two safety factors together. The resulting safety factor is somewhere between 1.67 and 2.0.

BA