I'm not a geotech guy but I have a question about a project I’m involved with about the skin friction of a H-pile in clay. According to the soils report, the soil profile, from existing ground and below, would include approximately 20 feet of sand then 45 feet of clay then 25 feet of sand. The blow count in the clay layer was commented as WH. I'm utilizing the method of Vijayvergiya and Focht (1972) to determine the average unit skin resistance for this layer (See attached formula). Since the blow count is 0, would it be appropriate to utilize this formula and if not what would be the appropriate formula or would you neglect skin friction for this clay layer. Any help would be most appreciated.
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Frictional (Skin) Resistance in Clay 1
- Thread starter Lake06
- Start date
You're right to be careful about whether to take some/any skin resistance from the clay. Given how soft the clay is, the first thing I would check is whether this material could still be settling under any recent imposed load (raising ground level for example). If this is the case, additional load will be exerted on the pile as the ground pulls the pile down. Are the blow counts 0 throughout the clay, or just in some areas?
Otherwise, look at the organic content of the clay and decide whether, in the longer term, you think this material will remain competent. Assuming a low shear strength of say 10kPa may not be unreasonable. Depending on how many piles you're driving and at what centres, installation may help consolidate the soil anyway.
Otherwise, look at the organic content of the clay and decide whether, in the longer term, you think this material will remain competent. Assuming a low shear strength of say 10kPa may not be unreasonable. Depending on how many piles you're driving and at what centres, installation may help consolidate the soil anyway.
- Thread starter
- #3
The site in this location was raised aprox 10 feet above the existing ground but this was at least 7+ years ago. No additional raising of the ground profile will occur due to construction. The blow counts are all 0 throughout the clay layer.
The soils report classified the clay layer as "Gray varved Clay, trace of sand lenses (CL)"
The soils report classified the clay layer as "Gray varved Clay, trace of sand lenses (CL)"
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- #4
motorbiketocrank
Geotechnical
It would be helpful to have more information such as what's being constructed, why H-piles are to be used, more background on the subsurface (ex. is it riverbed sediment deposits, etc).
I definitely wouldn't rely on any resistance/capacity from the clay. I think you really need to be worried about downdrag (negative skin friction). It takes very little settlement for this to occur. It could take many years for a 45' thick layer to fully consolidate. Plus, since total settlements are probably large, even a small % remaining could be a problem (i.e. if 1' of total settlement was anticipated due to the previous fill and 90% has occurred you'd have 0.1' still to go....which would definitely cause negative skin friction). Plus the clay might still be undergoing original depositional settlement. Or other factors such as any planned dewatering could cause the settlement. Would need more testing information on the clays to evaluate these sorts of things (consolidation tests with time curves, etc).
If you post more information on the planned development and subsurface (such as a boring log showing blow counts in the sands, etc) you might get better feedback on overall foundation recommendations. If you have negative skin friction from 65' of material (i.e. clay + overlying sand layer), that might completely rule out the viability of H-piles.
I definitely wouldn't rely on any resistance/capacity from the clay. I think you really need to be worried about downdrag (negative skin friction). It takes very little settlement for this to occur. It could take many years for a 45' thick layer to fully consolidate. Plus, since total settlements are probably large, even a small % remaining could be a problem (i.e. if 1' of total settlement was anticipated due to the previous fill and 90% has occurred you'd have 0.1' still to go....which would definitely cause negative skin friction). Plus the clay might still be undergoing original depositional settlement. Or other factors such as any planned dewatering could cause the settlement. Would need more testing information on the clays to evaluate these sorts of things (consolidation tests with time curves, etc).
If you post more information on the planned development and subsurface (such as a boring log showing blow counts in the sands, etc) you might get better feedback on overall foundation recommendations. If you have negative skin friction from 65' of material (i.e. clay + overlying sand layer), that might completely rule out the viability of H-piles.
Agree with geobdg and Rockmynci. That thick clay layer needs a lot of investigation before selection of a piling system. In addition to long term settlement and downdrag, I would question the ability of the clay layer to provide lateral support for piles, especially skinny steel piles.
- Thread starter
- #7
kieran1, the water level is 6 feet below the ground according to the soil borings.
I calculated the negative skin friction for this condition and found the downdrag force to be as great as the design load of 40 Kips. It appears the H pile (10x57) can handle an unbraced length for the entire length of the clay layer (45 ft), but the negative skin friction exceeds the allowable load now for a point bearing pile in the sand layer beyond the clay layer. Is there any kind of coating that can be applied to the pile to reduce the effects of the negitive skin friction, if so how can you calculate the reduction?
I calculated the negative skin friction for this condition and found the downdrag force to be as great as the design load of 40 Kips. It appears the H pile (10x57) can handle an unbraced length for the entire length of the clay layer (45 ft), but the negative skin friction exceeds the allowable load now for a point bearing pile in the sand layer beyond the clay layer. Is there any kind of coating that can be applied to the pile to reduce the effects of the negitive skin friction, if so how can you calculate the reduction?
Sounds like you need more embedment in the sand layer so that the nuetral plane is not located in settling soil. I suggest you read papers by Fellenius regarding negative skin friction and capacity. Sometimes negative skin friction isn't a bad thing. Google him and you will find many papers freely downloadable.
Lake06,
I've had that happen with a 60 foot augercast pile.
With the negative skin friction, the capacity was basically negative.
The pile worked fine at 80 feet due to a layer of 50+ blow count sand starting at about 40 or 50 feet down.
I would be leery of trying to use any kind of friction break.
Just embed the pile deeper.
I've had that happen with a 60 foot augercast pile.
With the negative skin friction, the capacity was basically negative.
The pile worked fine at 80 feet due to a layer of 50+ blow count sand starting at about 40 or 50 feet down.
I would be leery of trying to use any kind of friction break.
Just embed the pile deeper.
motorbiketocrank
Geotechnical
There are various ways of reducing NSF such as coating the piles with bitumen but they're not foolproof and you have to worry about damage of the coating during pile intstallation (and of course you wouldn't want to coat the portion of the pile that would be embedded in the lower sands).
Are you sure you need to use piles on this job? What about ground improvement techniques? And if using piles are you sure you want H-piles (where the large surface area makes the NSF problem worse)? For instance a pipe pile would have less surface area for NSF and would have increased end bearing.
I think you may have real problems with the H-pile approach. If your upper and lower sands were of similar density you'd have to drive the piles 20' into the lower sand just to offset the upper 20' of sands NSF (and more to offset the clay NSF and structural load). I realize your sand zones may have different densities and you also have some end bearing capacity but just trying to point out intuitively the problems you're faced with.
And beyond that I think the biggest problem may be that you can't drive the piles sufficiently to achieve the capacity you need. While the NSF pulls down on your piles once placed, the material will be working against you during installation (primarily the upper sand). Plus the longer the pile, the less stress will be transferred to the tip. Getting enough driving stress down 65'+ to penetrate the lower sand far enough to offset the eventual structural and NSF loads may not be possible. You'd need to run a wave equation analysis to get a feel for this.
Of course you might consider vibratory techniques for installation but that has it's own problems and doesn't guarantee success. And final thing - sounds like you might need to be installing 20' or more into the lower sands......which is about the limit of the soil profile you presented.....what's below that?
Are you sure you need to use piles on this job? What about ground improvement techniques? And if using piles are you sure you want H-piles (where the large surface area makes the NSF problem worse)? For instance a pipe pile would have less surface area for NSF and would have increased end bearing.
I think you may have real problems with the H-pile approach. If your upper and lower sands were of similar density you'd have to drive the piles 20' into the lower sand just to offset the upper 20' of sands NSF (and more to offset the clay NSF and structural load). I realize your sand zones may have different densities and you also have some end bearing capacity but just trying to point out intuitively the problems you're faced with.
And beyond that I think the biggest problem may be that you can't drive the piles sufficiently to achieve the capacity you need. While the NSF pulls down on your piles once placed, the material will be working against you during installation (primarily the upper sand). Plus the longer the pile, the less stress will be transferred to the tip. Getting enough driving stress down 65'+ to penetrate the lower sand far enough to offset the eventual structural and NSF loads may not be possible. You'd need to run a wave equation analysis to get a feel for this.
Of course you might consider vibratory techniques for installation but that has it's own problems and doesn't guarantee success. And final thing - sounds like you might need to be installing 20' or more into the lower sands......which is about the limit of the soil profile you presented.....what's below that?
- Thread starter
- #11
geobdg
The reason we are checking the H-Piles is because a previous engineer at my company did the design. However he no longer works here and I can't find any of his calculations. The building foundations are all designed so I wanted to see if the H-pile design could be saved. Assuming they work. I do have information beyond the soil profile I listed and the blow counts are in the range of 35-45 blows per foot.
The pile I am checking passes the driving formula I used. I have never used a wave equation before. From what I have heard its more accurate. Should I be using the wave equation instead? I see some building codes require it over 40 kips and that is about what my factored dead + live load is.
I have been reading a few downloaded papers by Fellenius. According to Fellenius the live load and the dragload do not combine and that two separate loading cases must be considered dead load plus dragload, but no live load and dead load and live load, but no dragload. He also states that the dragload must not be included in the consideration of the geotechnical capacity.
I guess I am confused, in my case I have a NSF almost equal to my design load, which is Dead + Live. Is this the correct combinations to be using when designing the pile. I am considering these end bearing piles and only using skin friction to check uplift, which is only about 14 Kips. I may be confusing NSF with dragload or are they the same thing. My experience with design of piles is limited.
Sorry for the delay in responding I got pulled off this project for a few days
The reason we are checking the H-Piles is because a previous engineer at my company did the design. However he no longer works here and I can't find any of his calculations. The building foundations are all designed so I wanted to see if the H-pile design could be saved. Assuming they work. I do have information beyond the soil profile I listed and the blow counts are in the range of 35-45 blows per foot.
The pile I am checking passes the driving formula I used. I have never used a wave equation before. From what I have heard its more accurate. Should I be using the wave equation instead? I see some building codes require it over 40 kips and that is about what my factored dead + live load is.
I have been reading a few downloaded papers by Fellenius. According to Fellenius the live load and the dragload do not combine and that two separate loading cases must be considered dead load plus dragload, but no live load and dead load and live load, but no dragload. He also states that the dragload must not be included in the consideration of the geotechnical capacity.
I guess I am confused, in my case I have a NSF almost equal to my design load, which is Dead + Live. Is this the correct combinations to be using when designing the pile. I am considering these end bearing piles and only using skin friction to check uplift, which is only about 14 Kips. I may be confusing NSF with dragload or are they the same thing. My experience with design of piles is limited.
Sorry for the delay in responding I got pulled off this project for a few days
motorbiketocrank
Geotechnical
Thanks for the background info. While I understand your predicament, I encourage you to choose the best solution, not just try to "save" one that may be inappropriate. If something goes wrong during construction or service life, people will be coming after you and they won't care about why H-piles seemed convenient at the time. Also, I'd recommend you involve an experienced geotech in the final design decisions. My instinct tells me driven piles are not likely the best design solution but I don't know enough about the project to say for sure.
Not sure what pile driving formulas you're using. Most I've seen are to correlate hammer blows to load capacity.....not to evaluate whether you can drive the pile through the material in the first place. While the wave equation can correlate the hammer blows to the load capacity, it can also perform a "drivability" analysis to see if you can install the pile to the depths required. It looks at the specific pile hammer, cushion type, pile dimensions, etc to figure out the stresses in the pile, etc. Basically, yes I would recommend this if you really are planning on using driven piles. If the contractor goes out there and can't drive the piles to the depths required, everyone is gonna have major problems. Remember though, with your situation, you are not going to be able to rely on blows as an indicator of capacity during installation. Plus, it will be difficult to perform a load test that will assure you meet your assumed capacity. (Since the upper soils will provide resistance during installation but can't be counted on for long term load support).
NSF and dragload are basically the same thing (NSF mechanism causes a downdrag load on the piles). You should design for the worst case of dead+live load and dead+NSF load. Since the live load is short term, the settling soils that are causing the NSF will provide a resistance to settlement from the live load (i.e. they are weighing down the pile but at the same time they are providing an equal capacity to resist the very short term live load).
About the "dragload must not be included in consideration of the geotechnical capacity", this is for evaluation of a plunging failure, not settlement. For example, say the soil is going to settle 2" and therefore cause NSF load. Ok, so then once the pile settles 2", there is no more NSF load. So although the pile has moved 2" (which is probably too much settlement), it has not experienced a plunging failure. Because the settlement can alleviate the NSF load (unlike with a structural load), NSF is not considered in plunging evaluation.
I'd question designing the H-piles based on end bearing (I can't believe you'll get a large capacity out of the end bearing in sand). Sometimes on rock or very dense materials this applies but not in most soils since the cross sectional area is small. Usually in soils they're designed for skin friction.
Not sure what pile driving formulas you're using. Most I've seen are to correlate hammer blows to load capacity.....not to evaluate whether you can drive the pile through the material in the first place. While the wave equation can correlate the hammer blows to the load capacity, it can also perform a "drivability" analysis to see if you can install the pile to the depths required. It looks at the specific pile hammer, cushion type, pile dimensions, etc to figure out the stresses in the pile, etc. Basically, yes I would recommend this if you really are planning on using driven piles. If the contractor goes out there and can't drive the piles to the depths required, everyone is gonna have major problems. Remember though, with your situation, you are not going to be able to rely on blows as an indicator of capacity during installation. Plus, it will be difficult to perform a load test that will assure you meet your assumed capacity. (Since the upper soils will provide resistance during installation but can't be counted on for long term load support).
NSF and dragload are basically the same thing (NSF mechanism causes a downdrag load on the piles). You should design for the worst case of dead+live load and dead+NSF load. Since the live load is short term, the settling soils that are causing the NSF will provide a resistance to settlement from the live load (i.e. they are weighing down the pile but at the same time they are providing an equal capacity to resist the very short term live load).
About the "dragload must not be included in consideration of the geotechnical capacity", this is for evaluation of a plunging failure, not settlement. For example, say the soil is going to settle 2" and therefore cause NSF load. Ok, so then once the pile settles 2", there is no more NSF load. So although the pile has moved 2" (which is probably too much settlement), it has not experienced a plunging failure. Because the settlement can alleviate the NSF load (unlike with a structural load), NSF is not considered in plunging evaluation.
I'd question designing the H-piles based on end bearing (I can't believe you'll get a large capacity out of the end bearing in sand). Sometimes on rock or very dense materials this applies but not in most soils since the cross sectional area is small. Usually in soils they're designed for skin friction.
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- #13
The driving formula I used is the Modified En Formula checks the driving stresses for a pile with a max hammer blow per inch of penetration between 12-14, as being the limit. What type of procedure or program would you recommend when you conduct a wave analysis.
Also what would you recommend as being a good reference for designing piles. I currently have Principles of Foundation Engineering 5E by Braja M. Das.
I will consult a geotech guy here but I was wondering what would the range of the internal friction angle for the clay layer with a 0 blow count be? Could it be in the range of 0-10?
Also what would you recommend as being a good reference for designing piles. I currently have Principles of Foundation Engineering 5E by Braja M. Das.
I will consult a geotech guy here but I was wondering what would the range of the internal friction angle for the clay layer with a 0 blow count be? Could it be in the range of 0-10?
motorbiketocrank
Geotechnical
Wave Equation software -
I'd recommend the 2 volume set by FHWA "Design and Construction of Driven Pile Foundations Reference Manual". For whatever reason they don't offer it in .pdf on their website as they do for many manuals but you can search around the web and find it (or order hard copies).
Typically for the clay you wouldn't use a friction angle but rather an "adhesion" value. This is usually correlated to the undrained shear strength (cohesion) of the clay. This is discussed in the FHWA manuals and other reference material.
I'd recommend the 2 volume set by FHWA "Design and Construction of Driven Pile Foundations Reference Manual". For whatever reason they don't offer it in .pdf on their website as they do for many manuals but you can search around the web and find it (or order hard copies).
Typically for the clay you wouldn't use a friction angle but rather an "adhesion" value. This is usually correlated to the undrained shear strength (cohesion) of the clay. This is discussed in the FHWA manuals and other reference material.
- Thread starter
- #15
motorbiketocrank
Geotechnical
No problem.
I really haven't used the driven program but am familiar with it. It serves a totally different purpose than WEAP. The driven program is used to calculate the capacity of the pile once it's in place. The WEAP program is to look at the stresses in the pile during installation (and relate them to capacity and driveability). Basically, driven is the design side of things and WEAP is the construction/constructability side of things.
For a typical project we'd first do the design based on hand calculations and/or computer programs. Then we'd check using WEAP to ensure the pile could be driven to the depths necessary without damage, etc. Then, in the contract specs, we'd require the contractor to do his own WEAP analysis based on the specific hammer, etc that he plans to use (which we would review). Then we'd also require a static load test and/or CAPWAP/PDA analysis to verify capacity.
I really haven't used the driven program but am familiar with it. It serves a totally different purpose than WEAP. The driven program is used to calculate the capacity of the pile once it's in place. The WEAP program is to look at the stresses in the pile during installation (and relate them to capacity and driveability). Basically, driven is the design side of things and WEAP is the construction/constructability side of things.
For a typical project we'd first do the design based on hand calculations and/or computer programs. Then we'd check using WEAP to ensure the pile could be driven to the depths necessary without damage, etc. Then, in the contract specs, we'd require the contractor to do his own WEAP analysis based on the specific hammer, etc that he plans to use (which we would review). Then we'd also require a static load test and/or CAPWAP/PDA analysis to verify capacity.
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