high reinforcement ratio for one story columns

[geo321]

hello. In reinforced concrete construction, if edge/corner columns are braced by a horizontal beam and the slab has been designed without relying on column stiffness, how can you limit the unbalanced moment absorbed by the column ? i dont assign release to the column since it is a RC construction and the connection slab/column will never be a pinned one. The column will always have to take some of the unbalanced moment. The problem is that, even for one story construction, the column reinforcement ratio ends up by being on the high side.

What do you do in general to encounter the above?
Would you arm your column accordingly or would you decrease your column capacity? Some firms check whether the section is cracked or uncrakced and assign modifiers accordingly to the column axis which is experiencing a high tension stress. Thanks in advance

 

[JAE]

Corner columns, especially at the top of one or multi-story buildings, typically do require more reinforcement than intuition might suggest.

Modeling your floor/roof system with the assumption of "pinned" columns is fine as long as you also include top steel at the perimeter and corners to compensate for the missing negative moment in your analysis in some rational way.

However, ignoring the connectivity between the floor/roof and your column and designing the column as though that bending isn't present is not conservative. It will be there.
So the column capacity should not be decreased.

You can, as you suggest, run an iterative analysis where you zero in on the proper effective moment of inertia of both floor and column due to cracking. This would be an attempt to be more accurate and less conservative in your design. ACI has suggested some presumptive values for Ie (0.7Ig for columns and 0.35Ig for beams, etc.) but for corner columns I'd be a bit more careful.

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[rapt]

What is actually bracing the floor?

Are there shear walls or some other bracing elements, or is the concrete frame taking all horizontal effects?

If the frame is providing the bracing, the Frame must be designed for the moments. You cannot assume pinned for the slab/beam design. Unless you design the columns as cantilevers fixed at the base, which is possible in some cases but not logical for a concrete framed building.

 

[KootK]

how can you limit the unbalanced moment absorbed by the column ?

Some ideas:

1) Make the beams stiffer.

2) Make the columns more flexible.

3) Adjust section properties for cracking as you've suggested. Tough to quantify this accurately.

4) Account for moment redistribution arising from creep, which can be significant and tends to reduce column moments usually. Very tough to quantify this accurately.

5) Design your gravity columns such that they form a flexural "hinge" at a particular level of moment. Basically, you design your columns such that axial load remains below the balanced point on the interaction diagram and tension bar yielding will precede concrete crushing. This makes some sense since, as you're finding, one story edge and corner columns tend to be loaded more like beams anyhow. Some conservatism is in order with regard to estimating the peak moment. That will tend to lead to pretty stocky column sizes.

6) Try and design something close to a hinge at the top of the column (Mesnager hinge etc). I've never seen this done in building construction. Too weird looking and I'm pretty skeptical of the performance anyhow.

What do you do in general to encounter the above?

Unless there's a serious congestion issue that I can't resolve some other way, I prefer to just reinforce for the moments. Some extra rebar in one floor worth of columns is unlikely to have much of an economic impact on the overall cost of the project. And columns are kind of important so I like 'em safe. When I do mess with column stiffnesses, I'm usually doing it to solve punching shear issues in flat plate and flat slab systems.

One major drawback of some of the fancier approaches is that they're... fancier. That stuff slows me down and eats up fee.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[JAE]

Just do KootK's implied "step 7)" and design for the moments. As he states - "That stuff slows me down and eats up fee."

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[geo321]

Thank you guys for the feedback. That is how i see it personally.
1- Stiffening the beam will stiff the support area which will lead to a moment increase.

2- Arming the columns with the required bars will lead to the fact that it should be developped. While the ACI requires, for tension development, a ldh (around 20db) and a standard hook, developping 20 or 25mm will necessitate a big slab. Rather, i go for 1.3xld and i bend the bars so the total length beyong the critical section is equal to 1.3ld (ACI does not treat this subject but i think it is a common practise to do it). decreasing the bar diameter will lead to a section reinforcement congestion.


3- I think a good way is to calculate the moment capacity which can be carried by the minimal top reinforcement located in the slab and compare it to the unbalanced moment in the column. If moment capacity is less, then column section reduction is totally fine and should not cause any problem which leads to a decrease in the column ratio. One thing is what is the tributary width of the column ? is it the same width used for punching shear calculations which is column width + 1.5 h from each column side?

Any comments ?

 

[KootK]

1) I don't know your specific situation but, in general, I would have expected the opposite.

2) Unless I'm misunderstanding your situation, what you need is a moment connection between your beams and your columns. It shouldn't affect your slab thickness and it take more than just development to get the job done. ACI 318 doesn't have much to say about these joints but ACI 352 definitely does. A lot of good stuff there.

3) Again, it sounds to me as though what you should be interested in is the moment capacity of your beams rather than your columns. I think it's unwise to attempt to cap your column design moment to the moment capacity of your beams or slab. You'll have to consider over-strength in the slab and rebar, degree of slab cracking, accidental T-beam action, and a bunch of other stuff that's going to be very difficult to conservatively quantify. If you're determined to design in a flexural hinge to limit column moments, I think it a much better strategy for that hinge to be in the column as I mentioned previously (#5)

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[geo321]

Sorry if i was not clear. Made a small hand sketch. As u can see, i have no problem at the connection beam-column (same spans, no moment transfered to column). However in the perpendicular direction i have the connection between slab and column which is experiencing a high moment.

 

[clarke1973]

The sketch shows zero moment at the base. If you are assuming a fixed connection at the top for monolithic construction, surely you should do the same for the base? I would assign some fixity anyway if not a full moment connection

 

[dik]

What I've done with parkades and other structures is used 70% of the column E value because it is higher on the stress-strain curve and E is less (if you're using a computer program that accounts for non-linearity of Young's modulus, skip that step), and also used a real anticipated live load for determining the maximum end span moments (for parkades 20psf).

The beams are designed for the end span moment as determined above and for the full live load. Makes a big difference in particular when you have a 60' span with a 16x36 Tbeam framing into a 14x30 wide column. As JAE noted, the top floors generally have the greatest percentage of steel.

Dik