HSS Col-Beams from four sides 3

[JoeBaseplate]

I want to verify the changes I want to make to this attached obsolete detail. I have a HSS col with WF beams coming in from four sides. I can get a thru plate for connection from two sides but I am not sure about the other two sides. Could I use welded plates (as marked up) which are not thru plates? Thanks.

 

[ToadJones]

connect-
Let me be clear, I was referring to the thickness to eliminate prying.

 

[connectegr]

Reason we avoid the welded hanger detail...

The flanges become very thick, when developed for bending. This requires "built" T's or pieces of much heavier sections to provide adequate flange thickness. The supporting member may also require reinforcement. This detail also requires overhead field welding, which is not for the typical iron worker.


I also apologize for leaving the thread topic.

http://www.FerrellEngineering.com

 

[connectegr]

Hokie
I have attached our standard for single plate connections. The shear capacities are pretty good for face welded plates. Note, that we provide a detail for the width/thickness requirements, to adequately account for punching shear and weld capacity. The thicknesses relative to wall width are pretty reasonable. If you are using "soda can" columns, then I agree that through plates or other connections should be used.


Thanks for getting back on the topic.

http://www.FerrellEngineering.com

 

[hokie66]

Thanks for that, connectegr. The capacities are surprising to me. I'll try to do a comparison with Australian practice.

 

[RFreund]

I'm a little late here but good thread discussion.

Just so I have this right -

The eccentric load on the WT creates a moment at the connection. The Tension component of the force couple pulls the WT away from the HSS which puts the flanges in flexure. The rotation of flange then causes the root of the fillet weld to be stressed, which is not ideal.
Correct?
Where does the prying occur?
See attached for an illustration of the questions.

Thanks.

EIT

 

[nutte]

RFreund, the connection you're talking about is a WT shear connection, where the WT acts the same way a shear tab does. Here the weld is generally considered to be in straight shear, with no eccentricity. The bolts are designed with the eccentricity.

The "prying" with WT hanger connections occurs when a tension force is applied to the web of the WT, and the flange of the WT is welded to an overhead support. In this case, as the flange flexes, the weld sees tensile stresses at the root, which can lead to the unzipping effect.

 

[connectegr]

RFreund
Your details correctly illustrate my concern with the WT shear connection. Current design examples for double angles and shear plates consider the beam web bolts to be in straight shear, but I am not aware of any design examples making the same consideration for the WT connection. As you have shown the eccentricity results in prying of the WT flange. However, nutte is correct and I am sure the others using this connection are not checking prying in the WT flange. The design example in the HSS manual does not address this concern. It actual limits the flange thickness to provide beam end rotation. The requirement for flexibility is also noted in the 13th Ed manual. The design example considers eccentricity in the bolts, bending in the WT web, but no prying on the WT flange. If the forces are large or the eccentricity is large, this tension at the top of the connection can be large. Although we frequently use knife-angles to embeds with the outstanding legs welded, we avoid the use of the WT shear connections. I prefer to consider the prying in the design, and expect that the supporting member will provide adequate ductility for end rotation.

Clearly a personal bias.

http://www.FerrellEngineering.com

 

[dhengr]

I don’t have the benefit of all the latest codes and manuals, so I’m hard pressed to see the very same code sections or manual figures you guys are sighting. However, for that end connection, on the simple beam that we are talking about here, basically assuming only a shear reaction, non end moment connection; you have the very same forces from the building and beam acting on the connection wether it is a WT or a single shear plate, they are just taken out differently into the HSS. And, as I see it they both have some drawbacks or conditions which are kinda hard to rationalize from the stress or weld standpoint.

For the WT as a hanger with the web vertical and with a hanging load, you will have a prying action causing tension at the root of the fillets, not a good condition; as Rfruend shows in his Sec. A-A, if the only welds are out at the flg. tips to the supporting member. But, you can start to alleviate this prying action by welding across the cut flg. on both ends of the WT to the supporting member. Then the problem I have trouble rationalizing is a fairly high weld stress within an inch or two on either side of the WT web because of the way the stiff web transmits the load the weld on the cut end of the flg. For the single pl. hanger you can certainly get enough double sided weld btwn. the single pl. and the supporting member, a good weld detail too, but now the starts and stops (terminations) of these welds have the potential of becoming highly stressed or stress raisers if there are any undercuts or craters, particularly if the load is not perfectly vert.

With the WT as a shear connection on the HSS you have the same considerations. I would weld across the top of the WT flg. to minimize the prying action on the vert. welds out at the flg. tips. The weld hard spot at the WT web becomes softer because the HSS face pl. is flexible, but this should still be looked at. The forces applied to WT by the building and the simple beam are the same as those applied to the single shear pl. We assume the rotation (some fixed end moment) will be taken care of by the slotted holes, but maybe not if the nuts are tightened. Otherwise, you still have a moment applied to the connection as a function of the shear and its eccentricity, Rfruend’s sketch pretty much shows it as I see it.

You can do the very same sketch for the single shear pl., and now any end moment (RF’s, Td = M) plus Pe = another M, are acting with their max. stress in the shear pl., from any moment in the shear pl., right at the start or stop (terminations) of the double vert. shear welds, at the top. And, without great care in welding and inspecting this is hardly a better condition than the prying on the WT.

I think Connecteng pretty will summarizes the problem, or quandary, by saying some worked out examples and some figures seem to ignore some of the forces or couples which might be acting on the connection. However, the idea that “The design example in the HSS manual does not address this concern. It actual limits the flange thickness to provide beam end rotation.” seems to fly in the face of what we are discussing here. It seems that almost invariably we make a second problem by trying to fix the first problem, and it is left to us..., good, thinking engineers to try to pick the least of the evils, and then hope they don’t slot the holes in the wrong direction, or some such.

 

[dhengr]

Every detail and connection we do should get this kind of basic scrutiny; in more detail as we are learning the ropes, and so we really do learn the ropes and learn to recognize good details; and with a somewhat lesser attention (or time required) as we gain experience and know when and where to look for problems. As our buildings and details get more and more complicated and are expected to do more, at higher stresses, this is the exact place that an engineer or firm dedicated to these details will really shine. They are doing this every day, and as Connecteng has demonstrated, they learn to know where and when to look more closely, or to make that extra calc., etc. At the same time, this thread would seem to indicate that there a plenty of bright engineers here, who are basically able to do this themselves, on their own designs.

The problem I have with the way Structural Engineering is going these days, is we are starting to produce engineers who are ‘one trick ponies.’ You do the positive moment on the beam, I’ll do the negative moment, he can do the connection bolts, and we’ll hire a firm to do the welding on the connection.... And, there isn’t anyone who has a handle on, or control over, the entire design. I don’t bother checking your moment, or design, or shop drawings, I claim I just didn’t see your error, or you just didn’t consider what I had done to affect the design; but now I do have someone to blame now, other than me. And, the attorneys and insurance providers love it.