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Post installed Holdown A.B. next to cold joint

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shacked

Structural
Aug 6, 2007
169
For residential construction where the seismic or wind uplift loads at the end of shear walls are relatively low I have a detail that I use when the edge distance governs the uplift.
Basically anything above 2.0kips ASD I specify this detail.

See pdf. Basically I add concrete to the side of the existing footing and dowels in order to transfer the uplift force in to the new concrete.

A plan checker is telling me that I can't do this because there is a cold joint too close to the edge and that ACI 318-14 section 17(anchorage to concrete) would treat this as a cold joint and it is un-acceptable.

I do not see an issue with this detail since the calculation already assumes cracked concrete! Is there something that I am missing? How would I persuade him to agree with me?

THanks
 
 https://files.engineering.com/getfile.aspx?folder=30dd83ce-47e2-40c8-b1c5-5e562c754d34&file=EXT_HD_DTL.pdf
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The reason I feel anchorage length of the anchor itself should begin at the new dowels is due to the first failure frustum you drew, what happens if that starts perfectly at the top of the hole drilled for the new dowels? Now your embedment begins at the bar (at best). I just don't see how those dowels are going to be fully developed on the far side of the potential failure cone in the long run. I guess I can get behind the detail in concept, but I think that once the numbers are ran, especially on a retrofit, that you won't properly have development of the reinforcing across the joint/failure cone, nor will you be able to develop the anchors appropriately either.

In my mind, the new dowels need to be fully developed beyond the potential breakout frustum assuming the new concrete isn't there at all. Is there enough concrete on the otherside of said failure line to fully develop the reinforcing? I have my doubts.
 
jayrod12 said:
The reason I feel anchorage length of the anchor itself should begin at the new dowels is due to the first failure frustum you drew, what happens if that starts perfectly at the top of the hole drilled for the new dowels?

If you dig into the theory for the failure modes for these kinds of anchors, you'll find that they assume that you lose a little, mini-frustum at the top of the embedment to account for something similar. So I don't feel that we need to be trying to anticipate that ourselves. And that's good because one could go altogether mad considering all of the possible failure frustums that might take place before reaching the one that we use for design.

jayrod12 said:
I just don't see how those dowels are going to be fully developed on the far side of the potential failure cone in the long run.

Why would they have to be developed on the far side of the failure cone? To my knowledge, nobody's suggested that this is a supplemental reinforcing scheme. Rather, I've been interpreting this as bare concrete anchorage with a cold joint tossed in to muddy the waters. By the book, the dowels should be fully developed into the existing concrete such that they can serve the shear friction function that I described but, as far as I know, that's the extent of it. The anchorage proper isn't relying on the dowels.

Frankly, my biggest concern is whether or not one can successfully drill 8"/16" into the concrete with 1.75" edge distance and not have the side of the concrete just spall off. But, then, I'm always impressed at what seems to be able to be accomplished by construction pros.

@OP: out of curiosity, is it accurate to assume that the "blob" wold be installed before the anchors are drilled?
 
I appreciate everyone's comments.

Kootk....good point an whether or not the new concrete is installed before the holes are drilled. That is something I haven't thought about. Now that you mention it I do believe the new conc should be installed before any new holes are drilled into the concrete. Now whether or not they follow these steps during construction is another issue.

Sandman: Unfortunately in this particular case I have two short shear walls, one 4.5ft wide and the other 4'-3" wide with a 9ft plate height. If I use 2 holdowns per end post then my shear wall width would be reduced by approximatelly 9" total. 3" for each holdown and 1.5" for a trimmer at each end of the wall. Therefore the uplift increases by approx 20% and with 1.75" edge dist that doesn't work.

This has really got me thinking about the history of concrete anchorage with respect to old appendix D and how accurate the calculated allowable uplift forces are compared to real lift test results.
 
I've got some pull test data that I'm not sure I can share outright, but I'll describe the test setup and results. I won't comment on the detailing used and the implications it had on the test results, or whether we can extrapolate from these results. Just wanted to share to demonstrate the efficacy of the approach.

Test Block: 4000 psi concrete, 18" wide x 24" thick x 30" long.
Anchorage: 5/8" diameter epoxy anchor with 8" embedment, centered on length of test block, 1-3/4" edge distance.
Section Enlargement: 6000 psi concrete, 6" wide x 12" thick x 16" long.

Enlargement reinforcing was as follows:
[ul]
[li](2)#4 bars spaced 4" o.c. (straddling the anchor) epoxied 8" into the test block (2" from the top) and hooked alternate directions in a horizontal plane into the enlargement. Hook extension was 9".[/li]
[li](2)#4 bars spaced 2-1/2" o.c. (centered on anchor) epoxied 4" into the test block (1-1/2" from the bottom of the enlargement) and hooked in a vertical plane into the enlargement. The hook turned up toward the top of the enlargement and then turned 3" back toward the test block directly on top of the horizontal bars from bullet point above. [/li]
[/ul]

Without the section enlargement, load at failure was about 20 kips. With enlargement, load at failure was about 31 kips. So a pretty significant benefit.

 
Baby... nuclear! Do we know anything about the ultimate nature of the failure with the enlargement in play?
 
Ultimate failure was by breakout. A large frustum typical of what you would expect developed in the test block along with a smaller underdeveloped frustum in the enlargement. The compression strut emanating from the anchor caused the enlargement to rotate about the bottom of the joint, causing that bottom corner to spall while creating a large separation at the top of the joint.
 
Thanks for sharing that info Deker. I figured that the failure would probably be somewhere in the range of 8 to 10kips, but not 20kips.

 
Gonna pull a little well intentioned malfeasance here. This seems to have run its course so OP probably won't mind.

@Decker & Sandman21: there's an interesting thread going on here that you guys should join in on. The use of multiple risk category classifications within different parts of single structures designed via PBD in the Los Angeles market.
 
I apologize if i missed it, but I don't suppose you can rely on anchor reinforcement? Others have mentioned treating it as development length... that is essentially included in chapter 17 (old D). If you can look to that, you will skip this bs and follow 17.2.3.4.5 to 17.4.2.9 and hence to chapter 25. Given the loads we are talking about, you wouldn't need much steel to exist. It might be enough to have a couple verticles known to exist, and be inside a couple perimetre bars. Even then, you have a bit of an issue with your offsets regarding those perimeter bars, requiring reduced cover (only by 1/8" with #3s). It does say "existing" in that detail...
 
I agree that you're stuck with the 1 3/4" edge distance for the Appendix D checks because you will not develop the full pull out cone. The cone will be interrupted at the face of the cold joint.

However, it appears that you may be able to create a load path for shear friction to develop via the horizontal bars across the cold joint.
 
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