Structural Steel- Moment Connections 2

[LoneStarEngineer]

When designing a steel moment frame, if the beam is designed to support the full gravity load as a simply supported beam and the connection is designed for moment due to wind only (instead of envelope combinations including Dead and Live), how does this affect the drift? Do we have to check the drift due to just the nominal wind load or can we can we check the drift from the envelope load combinations?

If we are taking the drift from the envelope load combinations, should we design the beam and moment connection for the envelope load combinations as well?

To put things into perspective, for a sample 2-story moment frame, I am seeing a moment of 160 k-ft (from ASD Envelope Load combinations) and the wind moment is just 33 k-ft. I am questioning whether I should design the beam and moment connection for 160 k-ft or 33 k-ft? (After already designing it to support the gravity load)

Appreciate your responses.
Thank you.

 

[E720]

Assuming that the Seismic Provisions are not in the scope of what you are doing, everything needs to be designed for the maximum forces from the envelope load combinations, even if it is a moment strictly due to gravity loads on a fixed connection. For gravity loads, the beam will see less moment when modeled as a fixed-fixed beam (wL^2/8 vs wL^2/12). It's the connection that is really going to take a hit. That is my understanding, I am happy to be corrected if I'm wrong.

If you don't want the connection to be able to rotate under lateral loads then you will also have to prevent it from rotating under gravity loads. You can't just fix it in one loading and pin it in another.

 

[KootK]

I think that the concept that you're looking for is "Flexible Moment Connections". Google is your friend. I personally dislike the method and would stick to designing your connections for the action envelopes that include fixed ended gravity effects.

 

[jayrod12]

I think that the concept that you're looking for is "Flexible Moment Connections". Google is your friend. I personally dislike the method and would stick to designing your connections for the action envelopes that include fixed ended gravity effects.

I'm surprised you even write that term down anymore after your last foray into FMCs

 

[LoneStarEngineer]

I think that the concept that you're looking for is "Flexible Moment Connections"

Exactly, I was trying to search in the 14th edition of AISC if they allow to do "Type 2 with Wind" design (from 9th edition) but couldn't find it. Has anyone on this forum designed flexible moment connections before? What approach did you take?

Thanks for all your answers.

 

[KootK]

Oh, you'd be surprised at the myriad of ways that I'm willing to sacrifice my integrity. I'll likely be bragging about the detail via a LinkedIn marketing post in the near future. It's a cool looking detail if you're a noob. Commenting functionality turned OFF naturally. Can't have you out there poking the bear for sport.

 

[LoneStarEngineer]

I personally dislike the method and would stick to designing your connections for the action envelopes that include fixed ended gravity effects.

I have HSS columns and the Envelope moment (including gravity and wind) is much higher than the HSS wall can take. I can't have diaphragm plate because it interrupts the sheathing on the exterior and the Arch would not be happy about it. Any thoughts in this scenario?

 

[KootK]

Has anyone on this forum designed flexible moment connections before? What approach did you take?

I did this last spring and took the approach outlined in the literature. You'll need the literature to do this properly. You know, to the extent that it's a "proper" thing in the first place. There's a truly excellent thread here on the subject as well. I believe that it was Lion06's initiative? My advice to you would be to get googling and return with any specific questions that you may have.

 

[southard2]

Personally I would only rely on the assumptions in Flexible Moment Connection design if

1) The moment connection was completed AFTER full dead loads were in place and
2) You are pretty sure that during a maximum wind event that the floor and roof live loads would be pretty small.

I have never used them as I simply do not trust this methodology. Save yourself some time and simply use beefed up moment connections.

John Southard, M.S., P.E.

https://www.pdhlibrary.com/

 

[KootK]

Any thoughts in this scenario?

Don't do it. I did the HSS thing during my foray into FMC and don't ever plan to repeat that. In the unlikely event that I ever get bamboozled into FMC again, it'll be WF columns, empire state style. Search the threads that I've started recently and you will find one on this exact topic. It's got a bunch of detail sketches from yours truly as well as some thoughtful advice from some of our best and brightest forum trolls.

 

[TehMightyEngineer]

I haven't designed a FMC but followed the various threads KootK indicated over the years. It was enough to convince me that FMC was inefficient at best and dubious in it's methodology at worst.

I'd review the references KootK made but begin planning on alternative framing. In my experience whenever I found I was overloading an HSS' wall it was much more efficient to switch to a wide flange.

Ian Riley, PE, SE
Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries

https://www.facebook.com/AmericanConcrete/

 

[LoneStarEngineer]

Thanks for all your answers guys! I got access to a journal paper by Louis Geschwindner- "Flexible Moment Connections for unbraced frames subject to lateral Forces".
I will be going through it over the weekend.

We have a senior engineer in the office who has used flexible moment connections throughout his 50 year career. I will have to find some hard argument to convince him. Anyways, I'll come back with any questions I may have.

Cheers and Go Astros

 

[KootK]

I will have to find some hard argument to convince him.

I'm your huckleberry. Those spiffy crossing graphs in the Geschwinder paper allow you to, effectively, use partially restrained moment connections without accurately knowing the moment-rotation characteristics of those connections. Pretty sweet. And legitimately clever. Unfortunately, the method still requires one to know that the moment connections provided are sufficiently ductile that they don't tear themselves apart in service.

As far as I know, the only way to establish sufficient ductility is through testing. And that, in my mind, means that you're stuck with the handful of WF column connection configurations that have been tested. As you've likely seen from my prior work here, I struggled mightily to come up with a ductile HSS moment frame connection. I basically resorted to trying to replicate some aspects of the yielding components present in the WF column versions. At the end of it all, my HSS columns ended up being enormous because the columns are just cantilevers for most intents and purposes. So I'd challenge your colleague to propose an HSS column connection that possesses adequate ductility and can be shown to possess that ductility. Without that, it's all just high minded jaw flapping.

On another note, FMC can be very inefficient from the perspective of design hours required. You have to disregard the leeward columns for various wind conditions which all but rules out software usage. This might not be too bad for rectilinear buildings with uniform bays in each direct but for more complex projects, it's an accounting nightmare.

 

[EngineerEIT]

Here are a couple of webinars which discuss these types of connections:

https://www.aisc.org/education/cont...ion-archives/flexible-moment-connections-n28/

https://www.youtube.com/watch?v=SdoZaCu7YLo

 

[bhiggins]

Here's a nice paper outlining the method of designing a wind-moment frame. They use end plate connections with undersized plates as a flexible moment connection. To calculate lateral drift, they suggest multiplying your calculated drift by 1.5 to estimate the drift of the flexible moment frame.

Here's a link:

https://www.steelconstruction.info/File:SCI_P263.pdf?internal_link

 

[LoneStarEngineer]

Thanks KootK and everyone for your responses.

EngineerEIT: I will check out those seminars. I had started watching the AISC webinar on these connections but need to finish it.

bhiggins: This paper references the British standards. I have read about this drift amplification factor in the paper by Geschwinder as well. From what I understand, this is suitable only for frames within a set of configurations (Only that has been tested) and cannot be universally applied.

I am getting an idea for why a majority of people here do not like flexible connections. Too many assumptions and too less factual data. Even if the connections work out, it would seriously bump up the steel sizes in order to control drift. I think I am gonna stick to fully loaded moment frames for now and not venture into the ambiguous world of FMC's.

 

[LoneStarEngineer]

Unfortunately, the method still requires one to know that the moment connections provided are sufficiently ductile that they don't tear themselves apart in service.

This was exactly what went through my head when reading the paper. The connection has to undergo permanent deformation before it is even ready to take in the wind moment. The method relies on knowing the real ductility of the connection which can be very difficult and far fetched when doing it from a theoretical perspective. If the gravity load moment is much higher than the wind load moment (which is true in my case), there is the uncertainty the connection would rupture and how much ductility can be achieved. And if the connection is more rigid than required, it puts more moment on the column.

Some of my colleagues were of the opinion that this would save a significant amount of money on our projects but it looks like an unfair exchange with the structural integrity of the project.

Thanks guys.