Trajectories for Principal Stresses 1

[releky]

This is the trajectories for principal stress for uniform load

http://imagizer.imageshack.us/v2/800x600q90/20/jyq8.jpg

I'm looking for illustration of trajectories of principal stress for concentrated load at midspan. Has anyone encountered this?

 

[releky]

The designer told me to add the carbon fiber asap as temporary solution as he realized plastic hinging may have deficient shear while awaiting how to analyze strategic tension trajectories undergirding stress-tensor states of field patterns and and phase-space topologies of the concrete beam-steel beam hybrid system where he has total lack of experience.

 

[rb1957]

if you're that worried about the building falling down, why wouldn't you support the beams with shoring and scaffolding ??


Quando Omni Flunkus Moritati

 

[AELLC]

Sorry, I don't have the time reading thru all this. Has anyone done any testing of the concrete itself?

 

[releky]

AELLC, in the article shared earlier, Vc of concentrated load at midspan is only one half. So concrete itself is not enough. In fact, it is ignored in the computation of plastic hinges. And considering there is deficient shear in the stirrups. This is one big headache. I took random sample of other designers work. They have deficient shear too. What happens is that many designers in Philippines use ETABS and STAADS, but they don't know the concept of shear and moment diagram. I asked them about this and they don't know what it even is. They just see the output table of the longitudinal bars. This is what happened to my designer who didn't bother to check the shear and moment diagram. So are half a dozen designs elsewhere. I will take more surveys. If most don't do this. It is possible over 50% of buildings in the Philippines have deficient shear in the beams.

rb1957, the building is already finished.. we can't use shoring and scalfolding forever.

The steel I-beam to be put under the concrete beam has to be very stiff or else it would deflect along with the concrete beam which means it can't stop the shear tension in the main beam. I wonder if there are very stiff 6 meter steel I-beam or a truss has to be specially made, and then the column metal plate has to bear the load. This is why composite action of them needed so the I-beam won't take the entire load and stress. Since only two I-beam is needed. It can be specially fabricated and being expensive is not a problem.

 

[AELLC]

What I meant is, frequently when there is distress in a structure, it is because the material strength is less than specified. Most often, the actual live loads are far less than Code-req'd live loads. (Exception: high-density filing systems), so even with a mistake in calculations, it is the deficiency in material strength that is the primary culprit.

In this beam, what is the ratio of actual dead load stress to live load stress (unfactored, working stresses?)

 

[n93s412]

AELLC - I believe the concern is with the stirrup spacing exceeding d/2 in a portion of the beam.

 

[rb1957]

"rb1957, the building is already finished.. we can't use shoring and scalfolding forever." ... understood, i was thinking about shoring as a temporary measure to protect the structure and the occupants.

Quando Omni Flunkus Moritati

 

[AELLC]

1) I skimmed thru all the posts above, and I still don't understand, is there actual cracking in the girder?

2) Has anyone actually checked span-depth ratio for compliance with ACI? Was the ratio exceeded, and an actual deflection calculation ever made, cracked section and all? You seem to be fixated on the mistake of calculation shear due to the concentrated load from the secondary beam, but to me, those girders are too shallow.

 

[AELLC]

n93,

You mean d/2 stirrup spacing was exceeded where the secondary beam occurs on the girder?

 

[releky]

rb1957, without seismic activity, no problem because the dead load of the beam and slab framing into the girder/secondary joint is only about 65 kN (14.6 kips). With live load of 4.8 kpa, the total live load framing into joint is about 87 kN (19.5 kips), but it's just office load which is lower. So the entire 1.2 DL + 1.6 LL total only about 215 kN (48 kips) and Vu is hence only 107 kN. The designer told me he assumed there are concrete partitions above beam so his Vu is 170 kN, but it's only 107 kN (24 kips).

But in seismic design, you have to subject the entire structure to cyclic loading (magnitude 7 earthquake) so the Vu of 107 kN become more than twice... 250 kN. Hence I-beam still needed to be put under concrete beam to resist solely earthquakes.
In the Philippines, we never use braced frame or shear walls... all are moment resisting frames but unfortunately.. construction is not of good quality.

 

[releky]

ALLEC, yes, d/2 is exceeded or d/1.5 at girder joint, some said only flexural cracks occur here.. do you believe shear can form right at midspan?

There is no actual cracking yet of the beams. No deflection yet because load is so low. The building beam shear satisfied only 1.2 DL + 1.6 LL but in seismic cyclic loading you have to double or multiply by 2 the entire 1.2 DL + 1.6LL and it is here where the building is not designed for earthquake as far as beam shear is concerned because the column sizes, bars, beam bars are designed for seismic load.. just the stirrups coming up short.

 

[AELLC]

OK I understand, but what about the designed depth of the beam and deflections when girder section is cracked? I get the feeling that the designers of this building are draftsmen who are trained to plug numbers into ETABS, not University graduates.

I can't get too worried about the shear at the secondary beams. But in a seismic event, I get worried about the shallowness of the girders and possible understrength of the concrete mix. Also, since you say construction is not good quality, what about proper placement of concrete, vibration while placing, etc? That is what worries me.

 

[releky]

AELLC, the designers of the building are civil engineering graduates operating ETABS headed by an structural leader (they have a dozen in the team). Right now. They are simultaneously designing 7 high-rise/mid-rise. See the company

http://rbsanchez.net/

I will talk with the team leader tomorrow or Monday again. I need to know what to ask. The depth of the girder is 500 mm (half meter). The girder with midspan concentrated load has shear to depth span of about 5.9. It's shallow. In the event of section cracks. ACI said the stirrups will hold it like truss. What can you comment about this depth thing during cracked section?

You asked earlier about the ratio of actual dead load stress to live load stress (unfactored, working stresses), the unfactored dead load is about 65 kN (14.6kips), superimposed dead load about 22 kN (4.94 kips), so total dead load + SD = 87 kN (19.5 kips). I plan to remove the entire superimposed dead load (tiles with 2 inches of mortar bed) to lower the dead load. The unfactored live load (typical office load) is 2.4kpa x 18 = 43.2 kN or if 4.8 kpa used is 86.4 kN but 4.8 kpa will be rare because the second floor will be office only (the building is just 2 storey). Hence, ratio of actual dead load stress to live load stress can be 1 or real life.. unfactored live load 1/2 (one half) less than dead load. What do you ask about the ratio?

About the concrete strength. It's 4000 psi and placement not good because space is tight. This is the reason we must put I-beam. Someone suggested that concrete beam - steel I-beam hybrid will work provided shear links/rods from the steel member to concrete element to be connected to have a monolithic response from both the elements. What can you say about this before I explore more ideas with the designers who haven't done such a thing before and since they are busy designing 7 highrise/midrise.. I have to say sense in the discussions because they don't have much time.

 

[releky]

AELLC, in addition to the above, I'd like to continue with the following facts:

http://imageshack.com/a/img23/9432/50os.jpg

Please see above layout of the floor. The designers don't do manual computations because they are designing so many buildings so they need sophisticated program like ETABS. They even forgot to manually compute so we are discussing the manual computations and comparing to their ETABS. Please comment on the following. I know the other formula to get tributary area of the slabs in each beam. But since we are computing the concentrated load in the girder/secondary joint, then the shaded area in the picture is the influence of dead & live load. I know there is contribution in the far side of the slab and beam as load varies along the spans but remember the other beams also contribute in supporting the shaded area load but the computation is similar if the entire shaded area is 100% of the load for estimation purpose. Isn't it this approximation is used?

Calculations of dead load
23.54 kN/m^3 (23.54 kN in a cubic meter of concrete)
Beam sections = 500mm depth by 300mm width
Slab thickness - 100mm (4")

Dead Load of secondary beams framing into joint

Beam framing into girder joint 3meters each so 23.54 x 0.5mx0.3m = 3.531x 6 = 21.186 kN

Dead load of the slabs.
23.54 kN/m^3 x 0.1m (thickness of slabs)=2.354 kN/m^2
2.354 x 18 square meters (the shaded area has 18 sq.m) = 42.372
So total dead load of the beam and slab in the girder is 63.558 kN. I didn't include the girder beam self weight because we are getting the concentrated load framing on the joint. Shear Vdl from the deadload is about 63.558/2 = 31.779 kN.

Live Load

assume 4.8 kpa maximum load 4.8 kPa x 18 sq.m - 86.4 kN live load
shear V = 86.4/2 = 43.2 kN

Total load
wu = 1.2 DL + 1.6 LL = 1.2 (63.558) + 1.6 (86.4) =76.2696 + 138.24 = 214.5 KN

or add SD load of tile floor = 23.54 x 0.05m (2") = 1.117 x 18 = 21.186
214.5kN + 21.186 kN = 235.686 Kn
but I plan to remove the entire 2" flooring mortar bed & tiles to reduce load so let's use 214.5kN (I won't allow any partition too in the shaded lines and ceiling weight below is so light because it is just light wooden board so let's ignore it for calculation sake esp since I round off the horizontal span from 5.5m to 6m)

From total wu of 214.5 kN

Vu = 214.5kN/2 = 107.2548 kN (24 kips) concentrated load framing into the girder midspan.

Vc and Vs of the concrete and stirrup is enough to handle 107.2548 kN.

So far, are the calculations above sound at least for 1.2 DL + 1.6 LL?

Of course if you take in seismic, the effect is like twice of 1.2 DL + 1.6 LL and I won't manually compute it. I'd just like to know for now if at least the building shear can support gravity load of 1.2 DL + 1.6 LL
Thanks.

 

[AELLC]

Releky

OK, the designer are engineers, but obviously they are very good at inputting into ETABS but lack practical knowledge.

I was curious about the ratio of live to dead because I was hoping the problem was not so bad (if the design live was very high) - I like to get an estimate of what the actual FoS is when something is realistically loaded (the actual live load is frequently much less).

Those secondary beams look much too shallow - if you exceed the ACI span/d ratio for beams, you have to justify by checking deflection of a cracked section.

When you deepen those secondary beams, the girders may still be overstressed since this is a moment frame. They look way too narrow. Girders with short spans relative to the beams they support are frequently wider than their overall depth.

 

[releky]

When you mentioned span/d ratio, do you mean the whole span over depth or shear span over depth? If the latter,
In the aci shear span/d ratio

a/d = 1 very deep beam
a/d = 1-2.5 deep beam
a/d = 2.5 6 shallow beam
a/d = very shallow beam

In the kani curve, very shallow beam would first fail in flexure before it fails in shear.
For shallow beam, it can fail in shear and flexure. My shear span/depth a/d of the girder is 5.9

If you are talking about clear span/depth ratio.. the secondary is 6 meters long, depth 0.5m so span/depth is 12, it's uniform load of the one way slab. There is 3 pcs of 20mm rebars grade 60 (60,000 psi or 414 MPA) bottom bars. Since the load is very low only... the end of short one way slab (see figure earlier), won't the bottom longitudinal bars be enough to resist deflection? I can put carbon fiber for flexure in the secondary beam if needed. But the moment of the secondary center dead load is only 18 kN. What do you think of this based on your experience so I talk about this to the designer... since if I don't mention anything, he won't say anything, so I have to question him... like he admit there is deficient shear after I question him about it.

About the girder. This is why we will put steel I-beam below it. Are you saying even for gravity load only, the beam depths are insufficient based on your experience or are they only deficient in seismic loading? What is your opinion about putting steel I-beam under the girder? Thanks for the tips.

 

[BAretired]

AELLC,

The secondary beams are 600mm deep. On a 6m span, L/d = 10 so I can't understand why you believe they are too shallow.

The girder is the same depth, namely 600mm deep. On a 5.5m span, it has L/d = 9.2 which, again seems okay to me.

I am not convinced that any remedial measures are required but if a steel beam is added under the existing girder, how will the steel and concrete beams be attached so they act in unison? How will the steel beam be moment connected to the existing concrete columns? How will the addition of a beam under the girder affect headroom? How will the removal of topping affect the fire separation?

It seems very strange to be reviewing design philosophy after the structure has been built, particularly when the design engineer doesn't seem to recognize or acknowledge that a design problem exists...or does he?

BA

 

[releky]

BAretired. No. the Beam depth is NOT 600mm deep but only 500mm deep (half meter) as I mentioned in all my posts, all have same size of 500 mm depth and 300 mm width. Would this make it shallow in your experience? On a 6 meter span, the L/d would be 12. Is this horrors?

Before the structure was built. I asked the designer why all the beam was only 500m deep. He said beam with 600mm depth is only for beam 7 meter span or more. For 6 meters or less, they use 500mm depth. But other engineers I asked said it should be 600mm deep. So is 500mm sufficient?

Worse case, I could categorize the second floor as residential unit to limit any big load.

 

[AELLC]

BA and Releky,

OK I see, the secondary beams are not so long. They just seemed so in the photograph showing the rebar.

So, BA, you are saying there maybe no problem, everything looks ballpark?

Its just that I am so conditioned to see girders that are wide and heavily reinforced. I really can't tell for sure unless I run actual numbers. Also, I prefer to use very old-fashioned techniques, and even so that I graduated from University in 1974 where they taught both the old working stress and ultimate strength concrete designs, I prefer to use working stress, hence my preference for deeper beams.

Releky, have you considered submitting the design for peer review by an engineer from New Zealand or from California or Nevada USA, someone well-trained in seismic?

 

[AELLC]

BA and Releky,

OK I see, the secondary beams are not so long. They just seemed so in the photograph showing the rebar.

So, BA, you are saying there maybe no problem, everything looks ballpark?

Its just that I am so conditioned to see girders that are wide and heavily reinforced. I really can't tell for sure unless I run actual numbers. Also, I prefer to use very old-fashioned techniques, and even so that I graduated from University in 1974 where they taught both the old working stress and ultimate strength concrete designs, I prefer to use working stress, hence my preference for deeper beams.

Releky, have you considered submitting the design for peer review by an engineer from New Zealand or from California or Nevada USA, someone well-trained in seismic?