Wall framing in conventional light-frame construction - IBC 2308.9 2

[sekwahrovert]

I have tried unsuccessfully to justify the stud heights in 2012 IBC Table 2308.9.1. I have attempted to apply worst-case loading as described in 2308.2. I am performing ASD analysis of members using all of the applicable load combinations. I am calculating member capacities using the 2012 NDS. Assuming stud grade Douglass Fir-Larch material, I can justify the strength of 2x6 framing at 16" and 24" o.c. but I cannot even come close to justifying 2x4 framing at either spacing. I have validated my spreadsheet results using RISA-3D, a commonly used commercially available analysis software, so if my input loads are correct, I am quite confident that my analysis is correct. I would appreciate any insights that might explain the problem I am seeing.

I appreciate all responses, but please don't tell me that "2x4's at 24" o.c. have been used where I live for years!" I completely understand the value of time-tested construction practices--that's simply not what my current question is. My problem is that I can't justify the stud heights from table 2308.9 by analysis. Section 2308 provides limitations on lateral and gravity loads which, if my analysis is correct, results in failure of 2x4s in combined compression and bending. As I mentioned, I have developed a spreadsheet solution which I have validated using RISA-3D. I'm hoping someone else has come to the same conclusion as I have and might have some ideas that might either help me reconcile my analysis and the IBC table or show why the studs in the table are ok even though a thorough analysis doesn't back this up.

My calculations are attached. The first page or so is the input/output summarized. The rest is more detailed (NDS adjustment factors, etc.).

 

[AELLC]

I don't think you can find 2x4 10' high in Stud Grade, use #2 instead.

With the wood stud compression and bending formula, a stress ratio of say, 400, doesn't mean it is overstressed by a factor of 400. The formula is very non-linear and tends to "explode" exponentially when you go above S.R.=1.0

 

[sekwahrovert]

AELLC, those are both very good points.

Regarding your first point, if I switch to #2, my 2x4 unity checks are 349% and 253%, so it still looks bad, but much better. This drastic change is related to your second point.

http://www.eng-tips.com/viewthread.cfm?qid=353341

Regarding your second point, the nonlinearity of the combined stress check did cross my mind, but I didn't really investigate how nonlinear it was. It is perhaps useful to look at the fb/Fb' and fc/Fc' checks to see how bad things are looking. In this case, now that I've switched to #2 DF-L, my worst-case fb/Fb' ratios are 89% and 59% for columns 1 and 2 (LC: D+0.6W), and my worst-case fc/Fc' ratios are 99% and 75% (LC: D+S). These results look much more encouraging. I'll have to see how much I have to reduce my loads before the combined stress unity checks come out less than unity. I'll post again when I have an answer to that question.

Thanks a lot, AELLC!

 

[wannabeSE]

A couple quick things (I only looked at the 2x4 @ 24"OC and I never use the convention framing provisions)

1. The IBC probably assumes the stud is fully braced. This will help the Fc
2. 35' mean roof height for wind is a too high for a single story building
3. for C&C wind load, you can use effective wind are of span x 1/3 span (or 33 sqft for 10' span)
I don't these thing will make it good, but it will look a lot better.

 

[sekwahrovert]

If we look only at bending due to wind load in the first column (2x4 #2 @ 24" o.c.), the bending stress ratio is 92%. If I start to load up the 2x4 with vertical dead load, introducing axial stress, the combined stress check goes over 100% when I place any more than 161 lbs on the member. That's not much. The combined dead load of the roof and wall is 744 lbs, resulting in failure in combined axial and bending. NDS Eq. 3.9-3 yields a value of 3.61. This is all shown in the attached PDF.

 

[sekwahrovert]

wannabeSE, thanks for commenting.

"1. The IBC probably assumes the stud is fully braced. This will help the Fc"
I am assuming fully braced for weak-axis buckling. Note the value of 1 ft in my calculations. Altering this value has no effect on my calculations until I enter a value higher than about 4 ft. If you're referring to compression edge unbraced length (for lateral torsional buckling), you may have a point. My intent was to account for garage walls which are sometimes unsheathed on the inside, but from what I have seen, are always blocked somewhere around mid-span at sheathing panel joints. If I make this value small, things look better, like you suggested, but still not good. Unfortunately, I don't think I can justify the smaller unbraced compression-edge length for a wall with no interior sheathing because I have to consider negative wind pressure, which is the larger of the two pressures listed in ASCE 7-10 Figure 30.5-1. Perhaps I'll just have to assume that all walls are sheathed on the inside and decide if we want to list this as a requirement in our residential designs.

"2. 35' mean roof height for wind is a too high for a single story building"
You're absolutely right.

"3. for C&C wind load, you can use effective wind are of span x 1/3 span (or 33 sqft for 10' span)"
Thanks, I'd never noticed the definition of "effective wind area" before.

My latest calculations are attached. Note that if I remove all live and snow loads, 2x4 studs @ 24" o.c. are now just slightly overstressed. NDS Eq. 3.9-3 yields a value of 1.07. I'm not saying it's ok to remove live and snow load from my analysis, but perhaps the conventional light-frame construction section of the IBC recognizes the small probability that design wind events will occur simultaneously with full live and/or snow loads. This is supposed to be covered by the load combinations I'm using, but perhaps the load combinations are more conservative than they need to be.

 

[woodman88]

Per my 2012 IBC
"2308.2 Limitations.
Buildings are permitted to be constructed in accordance with the provisions of conventional light-frame construction, subject to the following limitations, and to further limitations of Sections 2308.11 and 2308.12.
.....
3.1. Average dead loads shall not exceed 15 psf (718 N/m2) for combined roof and ceiling, exterior walls, floors and partitions.

Exceptions:
....
4. Vasd as determined in accordance with Section 1609.3.1 shall not exceed 100 miles per hour (mph) (44 m/s) (3-second gust). ..."

So Vasd = 130 mph wind is not allowed.
Will have to look closer at your other values and calculations.

Garth Dreger PE - AZ Phoenix area
As EOR's we should take the responsibility to design our structures to support the components we allow in our design per that industry standards.

 

[sekwahrovert]

woodman88,

I've attempted to implement the loading you described in my dead loads:
- 15 psf roof and floor dead loads applied over assumed roof and floor tributaries, 15 psf wall dead loads applied over half the height of the stud.
- 130 mph is the design wind speed per ASCE 7-10. Converted to Vasd, it comes to 100.7 mph, so we are actually in agreement.

Thanks for your comment.

 

[woodman88]

One thing you should be aware of is the 2012 IBC Table 2308.9.1 is the same as the 1988 UBC Table 25-R-3.
The 1988 UBC was under the old values and equations of the NDS. The wind C&C for walls @ 100mph would be 26psf(1.2@0'-20' or 1.3@20'-40')(walls at 1.1 and at discontinuities 2.0). I believe that the table heights could not be checked by the NDS at that time also. But I could be remembering this wrong.

Garth Dreger PE - AZ Phoenix area
As EOR's we should take the responsibility to design our structures to support the components we allow in our design per that industry standards.