Wind Loading of Unreinforced Masonry 1

[DemolitionMan03]

I'm evaluating a 1953 unreinforced masonry wall for bending due to wind load, and I'm having trouble understanding how the ASD Masonry Code (TMS 402-08/ACI 530-08/ASCE 5-08) handles the situation. The wall is non load-bearing structural clay tile, and it appears to be braced only at the top and bottom of the 25-ft height. I calculate a bending stress of 241 psi due to the C&C wind loading.

What do I use for the allowable bending stress: 1/3 * f'm (per Equation 2-17) or the value from Table 2.2.3.2? In my case, the first gives me 333 psi and the second gives me 38 psi. I went forward with the 333 psi and evaluated using the unity equation for flexure and axial compression (self-weight only here), and the wall is barely adequate (0.998). If I use the 38 psi, the wall (that has been standing just fine for 59 years) is 634% stressed.

What am I missing here? Do I not need to worry about the tensile stress (38 psi)? Any code clarifications, resources, masonry-for-dummies explanations, etc. would be appreciated.

BTW, this is my first post on the forum. I've check out the web site occasionally in the past, but recently switched jobs and am finding the site very helpful, so I thought I'd join up. I look forward to hearing what y'all think.

 

[dik]

How thick is the wall? In 1953, the building code was likely prescriptive as well as ASD based. Does your code have provisions for renovations that permits the use of the code in effect at the time the building was constructed (possibly with exceptions)?

Dik

 

[hokie66]

In 1953, most masonry walls were "designed" simply by empirical height/thickness ratio, without much consideration for lateral loading.

 

[msquared48]

I've seen a lot of older CMU structures with no cells grouted, maybe some horizontal K-web in the joints, and mostly old military structures. Some I dreaded entering to inspect. How they are still standing, I have absolutely no idea.

Mike McCann
MMC Engineering

http://mmcengineering.tripod.com

 

[DemolitionMan03]

Wow! Thanks for the quick replies, folks.

I probably should have provided more information on the evaluation. The building is a military building (good call, msquared48), a warehouse that was converted into a classroom building by adding suspended ceilings, HVAC fan coil units, and an open-web joist system to support it all. Some interesting problems there, but that's another thread. The base really wants to tear the thing down, but to do that they need money to fund the demolition and much more money to build a new facility for the classrooms. We are putting together an accurate structural as-built and evaluating what it would take to bring it up to current military code (UFC 3-301-01, which adds a few exceptions to the 2009 IBC, which of course points to ASCE 7-05, the MSJC, etc.). On one hand they want us to say it is awful and justifies the top priority for funding in the tight DoD budget. On the other hand, they are probably stuck with it for another 5-10 years, so they want to ensure it is as safe as they can.

I don't doubt it was originally designed on a more prescriptive basis, with little regard for lateral loading. And it is doing remarkably well for such an old structure that hasn't gotten a lot of TLC - a lot of overdesign and extra (cheap) materials and labor went into to those post WWII buildings.

Anyway, I still need to evaluate it based on current codes and methods. It would help the case for demolition to say these masonry walls are 600+% overstressed, but I don't think that passes the sniff test for a building that has been up for 59 years. The wall is 8" nominal thickness (7 5/8" actual). I can post all sorts of other values I've calculated and assumed ,but I think I'm just not understanding something about the masonry code here.

 

[concretemasonry]

Have you looked at the empirical design portion of ACI 530?

That ACI section for 530 is based on historical performance of unreinforced masonry as a guide to compare the older structure with the current 530 specification and code. That portion of the specifications does allow for 2-way performance of walls that are supported by other structural members. - Read the basis for the empirical section and do not try to apply the assumption from the other design/analysis sections oy you will find would find yourself wearing several layers of conflicting "belt and suspenders" garments.

It sort of is similar to the old story that an aeronautical engineer said a bumble bee cannot fly until he was chased by one. - Proven by performance without a high degree detailed assumptions and calculation.

Dick

Engineer and international traveler interested in construction techniques, problems and proper design.

 

[hokie66]

A 25' high, unreinforced, 8" block wall would never have complied with any code, empirical or otherwise. 18 x nominal wall thickness would have been the most typical requirement for an exterior bearing wall, and that would have restricted the height to 12'.

 

[dik]

Are there pilasters in the wall? As Hokie noted, 25' is awfully high... maybe 24" deep or with structural steel at 12' centres?

Dik

 

[RobertHale]

To answer your first question: the 1/3*f'm is the limit on the compressive stress from bending. This value is typically used when you are checking the stresses in a reinforced masonry section. The Table stresses give the allowable tension limits on the mortar joints. For a typical unreinforced section this will control as the tensile stresses will never be sufficient to allow the compressive element to reach the limit. If you have significant axial stress it is more likely that the compressive bending stress will approach the limit, but then the stability factor in the axial compressive stress limit will become significant. I do not have the code in front of me, but also check the section of the code the limits are in. I thought that the 1/3f'm only applied to reinforced masonry design. I hope this helps.

 

[DemolitionMan03]

No pilasters in the wall. The warehouse has a structural steel frame, columns spaced at 33 ft and (best I can tell) the masonry wall is only anchored to them at the top.

Thanks for the explanation, Robert - that makes sense to me. I guess I would word the code differently to specify that 1/3 * f'm is the compressive limit. As it is, the say the unity equation (and by extension 1/3 * f'm) applies to "members subjected to axial compression, flexure, or to combined axial compression and flexure". As you said, I don't see how the compression would ever control for flexure-only loading, so why have that provision? BTW, this is all out of section 2.2 (Unreinforced Masonry) so it should apply.

Thanks for the pointer on the Empirical Design chapter of the code, Dick. I'm checking it out now.

 

[JedClampett]

To answer the question "How is it still standing?" maybe the materials were better than assumed, maybe there's alternate load paths, maybe it's never seen full wind. 90 mph windstorms are pretty rare, especially inland. If you want to nurse the building to demolition, maybe you can add vertical steel beams. Or calculate the maximum span it could take and brace the wall at that point back to the roof. If you have a pretty firm demo date, maybe you can use a shorter recurrence for the wind (10 year). But be forewarned that unless the building starts to fail, there's very little urgency to tear it down.
I once had a professor who load tested a concrete floor to failure of a building that was going to be demolished. It tested to about 100 psf, which was the design value. When they tore it out, they discovered there was no reinforcing in the floor. He said that the only mechanism that could explain it standing was inertia.

 

[ishvaaag]

... and so the students learnt how much can be done with just words, Jed ...

 

[concretemasonry]

When it comes to masonry the real reason masonry buildings work and stand well despite what the assumed numbers say, is that the masonry units for the last 60 years or so are far stronger the assumed or minimum strength.

The ASTM C90 specifications are not considerably different than the old requirements in the 1920's and 30's. Manufacturing equipment and methods have changed significantly, but the standard for the units have not really changed. For working stress or other engineered masonry design methods on key number that everything keys off of is the f'm, or masonry strength. All too often people mistakenly use a low assumed (often minimum) in an analysis. Over 40 years ago, most producers routinely shot for 30% over minimums and now, it is more economical to make units 40% over the minimums to make a unit a contractor will buy and lay.

The zero slump concrete in masonry units is unique in that is very easy to make high strengths by just adding water, but the appearance suffers and they can be unsalable. In the last 40 or 50 years, the material handling and curing has become very sophisticated and the higher strengths (beyond ASTM standards or most engineers requirements) just happens because of the compaction and density that is produced.

Even the block made in a primitive way in many countries - molded using variable aggregates, spraying with water and covering for a day or so still easily exceed the ASTM and most engineers requirements.

The ASTM standards have been very slow to respond and producers are not interested in raising the minimum requirements since the proper use of ACI 530 that are performance driven if the engineers bothers to find out what is readily available.As I was an ASTM member, it took me about 10 years to be a voting member on ASTM C90, C270, etc and over 20 years later, the standards have not changed appreciably because it is essentially a consensus standard with a mixed membership.

Dick

Engineer and international traveler interested in construction techniques, problems and proper design.

 

[hokie66]

Dick, your answer may be correct for typical bearing walls loaded in compression, but the OP's question was about what is essentially a pure bending element. Very little to do with the quality of the blocks.

 

[concretemasonry]

hokie -

I was suggesting an approach to look at a means to provide a "base-line" for the realistic strength of the structure using continuity and available and reliable connection points, even if it gets to treating the walls as a flat plate with different connections/continuity. It is a real challenge in engineering and understanding of the behavior of materials to get to a baseline to work off of for further suggestions.

When it comes to shear and tension, most allowables are geared toward the masonry strength and the mortar type, which ASTM C270 suggests using the lowest compressive strength possible to provide the additional benefits beyond compressive strength for the wall to function as it is. I can remember when there was a 50% decrease in many allowables for flexure, for non-inspected structures, but the inspection process was not even accurately defined. It is very rare for a compression failure in concrete masonry, but common in the interaction with the structure's behavior.

Some of the worst disasters I've seen were near the Northridge,CA earthquake where portions on the buildings were arbitrarily over built (reinforcement and grout) and changed the building response and attracted more loads than assumed in the design process. Unfortunately, the damage was done before the loads could be distributed as initially assumed and the structure (adjacent to a hospital) failed. Northridge did have a huge vertical acceleration, but the lateral forces were still very significant.

This subject building appears to be a big box with repetitive details and connections, so base line "comfort zone" can be detwermined to based recommendations on.

I am a "loose" engineer tempered with a lot of strange/varied experiences and code/standards writing in many places. - That is the luxury of being older and a civil/structural engineers. - It is a great profession.

Dick

Engineer and international traveler interested in construction techniques, problems and proper design.

 

[hokie66]

Well, forgive me if I still think your comments are irrelevant to the OP's problem. He has 8" thick unreinforced masonry walls, 25' high, and braced only at the top at 33' centres. Nothing about the quality of the masonry units, or indeed the mortar, is going to help with that situation.

 

[JAE]

I've run a 25 ft. tall wall with 19 psf wind load - no grout, no reinforcement, running bond - using an f'm = 3000 psi.
This uses Clay block 7 5/8" thick.

This is with the NCMA program.

No other loads except self-weight.

As can be seen in the diagram - the wall isn't even close to being adequate with those parameters.

 

[DemolitionMan03]

Thanks for the replies and insight, everyone. I think we can all agree that this wall doesn't meet current code, but it is probably still standing because of the safety factors in material, design, and loading. However, I could use some help figuring out the solution.

I'm calculating the bending stress parallel to the bed joint as fbparallel = Mparallel / Swall. And the moment parallel to the bed joints as Mparallel = wL^2/8, where

w (plf) = wind pressure (psf) * 1 ft strip
L (ft) = height (ft)

Then I think the solution is to add girts at the third-points of the wall height (designed with sufficient anchorage to resist positive and negative wind pressures). The girts bring the unbraced height down to ~8 ft and reduce Mparallel (and therefore fbparallel) by a factor of 9. Any problems with that analysis or solution?

What I can't figure out is how to calculate is Mnormal. If I follow the approach above and turn everything 90 degrees, then L = the span (16 ft in this case), and that doesn't reduce unless I add additional vertical members, but it really doesn't seem like that shouldn't be necessary. Does reducing the unbraced height reduce Mnormal by a factor of 3 somehow? What am I missing?

 

[dhengr]

I’ve used the clay block or oversized hollow brick that I think JAE is talking about in the past for multi-story bearing walls. That’s quite a different animal than the old structural clay tile, infill wall panel, from 60 years ago, in both strength and unit wall thicknesses. The OP’er. doesn’t give much info. in the way of actual tile strength, dimensions, the various important details, etc. He claims primary wall support at the foundation/fl. slab and at stl. eave framing at 25' high; and then stl. frames every 33', but no info. on support there, or control joints, etc.. So the wall is probably spanning primarily in the vertical direction, and the stl. framing seems to be able to handle this, if there is no evidence to the contrary.

A couple possible solutions, other than the girts at third points might be: every 6 or 8', whatever turns out to be appropriate, saw cut the inner shell vertically so you can open a core or two, install rebars and connections t&b and grout the void full. There are also some post-tensioning systems available where you open up the inner shell at the t&b so you can work in the wall cavity. You epoxy (or some such) a connector into the foundation, snake the tensioning bar down through the cores, apply a cap plate at the top, tension the rod, and tighten a nut down on the cap plate. You could apply enough post-tensioning compression to the tile wall to over come the bending tension at the bed joints.

​

 

[hokie66]

DemolitionMan,
Now I am confused. Bending parallel to bed joints is horizontal bending. Bending perpendicular to bed joints is vertical bending, in which case I think the allowable tensile flexural stress should be zero for unreinforced masonry.

Where did the 16' come from? You said the frames were at 33' centres.