Wood Laminated Posts - does this work?

[allisch]

There is a laminated post company which is manufacturing laminated 2x wood posts for pole barn type structures. We are in a 90 mph wind zone.
They seem to show that they can use a (4)ply 2x6 post for wall heights to 24' with posts at 8' oc. (#1SYP material)
30 psf LL, 7 psf DL, 90 mph, negligable earthquake
They specified a maximum building size of 32' width.

I see the design wind load for the "non-corner" part of the building as being 13.6 psf suction (per IBC Table 1609.6.2.1(2)), which gives approx 110 plf.
If the post is properly embedded in designed footing, then I would use a fixed base and pinned top support.
Max Moment = wl^2/8 where L = the sidewall ht=24'
M=7850 ftlbs applied due to wind


According to my NDS 2005, I see #1SYP 2x6's as Fb=1650, E=1700, Fc=1750
S=33in^3 if the post is turned the strong direction to resist wind.
(Assuming the post is laterally braced (weak axis) 2' on center by girts, but only braced at the floor and ceiling level (24') in the strong axis.
I see the Moment capacity =4540 ftlbs at 100% or 6040 ftlbs at 133% wind stress.
So the wind stress in bending is 173%

So I think the post fails without even adding the gravity loads.
I have that the post has a design axial load of 11.4k.
If the building is framed with trusses (and posts only on the perimeter) in the 32' direction, with 30 psf LL, 7 psf DL, and 90 mph winds, the applied axial load under combined wind + dead + live is approx 5.3k.
So the axial load is using 47% of the design stress.
(I use column Ke=0.8 for fixed base posts-pinned at the top)

I missing something? Do you see any way that they can make this engineering work?
I don't see how they can have engineering to make this work.

 

[DaveAtkins]

A couple suggestions--

The load combinations should be:
* D + L
* D + W
* D + 0.75L + 0.75 W

The load duration factor for load combinations involving wind is 1.6, not 1.33.

Did you use the size factor and the repetitive member factor? They apply in this case.

DaveAtkins

 

[allisch]

Dave-
Thanks-

It looks to me like the repetitive member factor applies only for members spaced not more than 24"oc

As to the 1.6 factor- Are you referring to the Load Factor of 1.6 for the LRFD analysis? or are you saying that the wood is allowed 160% stress under wind loadings?

 

[Lion06]

Just look at the NDS. The load duration factor for wind is 1.6 (this is the Cd factor)

 

[spats]

I agree that it doesn't work. I just ran your column in Woodworks Sizer, assuming a 6x6 post (close to what you've got), #1 SYP, fixed bottom & pinned top, continuously braced in it's width direction, unbraced in it's depth direction, with the wind and axial load you describe. I input the axial as all live load to get the higher load duration factor. Axial load eccentricity = 0. My Sizer version uses NDS 2001. It is giving me a combined stress ratio of 2.85 using IBC load combinations.

Also, why are you assuming pinned at the top? Is the building braced?

 

[allisch]

OK I see the 1.6 Wind factor now-
Sorry- I had a Brain fart The 1.33 was a carry over from the old codes.

Table 4B Cf notes mentions that size factors have already been included in the table values, but mentions that if we got to be more than 8" wide (which we are not) that additional factors may apply.

 

[JKW05]

Careful... don't confuse the "Load Duration Factor, CD(NSD 2.3.2) used with ASD and the "Time Effect Factor", lamda (NSD 2.3.7) used with LRFD.

 

[allisch]

Spats- Thanks for the info.
I am assuming that the building shell and roof acts as a diaphragm

 

[OldPaperMaker]

Allisch,
If this is a pole building and the posts are cantilevered out of the ground then the bending moment is wL squared over 2, not over 8.
Also, when calculating the Column Stability Factor,Cp remember that the K value should be 2.1 per Appendix G of the NDS when you calculate FcE.

 

[allisch]

Old Paper maker-
wl^2/2 would mean that they have NO lateral support at the top.
I am giving them the benefit of the doubt that they can get the diaphragm strength from the sidewalls and roof to have the top as a pinned connection.

 

[allisch]

old paper maker-
but you do bring up a good point-
they also may have problems getting the roof to take the diaphragm loads to the sidewalls.
But that is a whole different can of worms.

 

[csd72]

I would assume that they are treating it as category 1 as this includes agricultural buildings. Therefore I = 0.87

 

[allisch]

No- these will have commercial uses

 

[JAE]

allisch,

We have experienced almost the exact situation you describe. These pole barns have been put up for years and generally stand fine - mostly due to the fact that your design 90 mph wind doesn't ever actually occur except in very severe storms and then many of those are tornados and the failure is attributed to that.

In recent years, these pole barn suppliers have encroached into areas where the IBC governs, and in many cases, the building requires an engineer.

We've been told that our problems getting their columns to work are crazy in that they've built hundreds of these over the years with "no problems".

The pole barns are generally designed with pinned bases and fixed tops...in that they use a knee brace of sorts (or some sort of rigid steel elbow) at the top in an attempt to create a rigid bent (2 columns and a truss working together to resist the wind).

So the columns are sway columns (k = 2) with a rotationally fixed top and a pinned bottom. When the MWFRS forces are applied, both columns and trusses resist.

The other problem we've seen is, despite the wind bent concept, the trusses are never really designed for these end moments.

When they essentially are faced with having to design per code - it just doesn't work. Plain and simple.

 

[msquared48]

The only way this has a chance to work is if the columns can span the wind load from the foundation to the roof diaphragm, the roof diaphragm is sheathed as a structural diaphragm, and the endwalls have rated shear walls with holddowns (or the multiple stud columns).

I, too, have designed too many of these over the years and have heard the same comments from owners/contractors.

24' is a ridiculous eave height for effectively a rough cut 6X6 column. May as well have rubber columns.

Looking more at 6X12 or 8X12 columns without the roof diaphragm and end shear walls. Have to use a grade of lumber that is pressure-treatable too, and that means a lower allowable stress grade.

The above comments pertain to the forces parallel to the roof trusses. The endwall columns would have to span vertically to the roof diaphragm, and the roof diaphragm be capable of distributing the lateral load to the long wall columns, that may be able to take the load collectively.

Mike McCann
McCann Engineering

 

[ctcray]

I too have had the problem of proving a pole building works. I specifically was dealing with this "product"

http://www.socketsys.com/index.php?id=tZa&goog=snow+load+tables

It alleges to have engineering behind it, but I couldn't justify it.

Anyone see this before?

 

[msquared48]

Have not seen this particular one, but there is a design similar in my area that uses pre-manuf metal brackets to insert the memmbers into. This one appears to have large bay spacings that could be questionable for the wind loads we have here. Base connection is not readily available from the advertisement. Knee braces for the 14' eave heights as expected for the members used. Would have to be thoroughly checked.

Mike McCann
McCann Engineering

 

[DaveAtkins]

OK, I actually sat down and put this into the spreadsheet I developed for wood stud design, and I find that it is marginally acceptable. For DL + WL, I get 0.879 from the interaction equation, and for DL + 0.75LL + 0.75WL, I get 1.14 from the interaction equation (a little high for my taste, but not unsafe by any means).

But I think the deflection would be quite large.

DaveAtkins

 

[allisch]

Dave-
Thanks-Can you give me an idea of how you get this to work?
Even with the 1.6 factor, which I goofed on and used 1.33 in the original post, I think it fails with W alone.

What is the design moment?
Are you using 13.6 psf suction? = 109 plf
Are you using 24' as the post span?
Are you using wl^2/8?
What Fb are you using?
Are you using S=33?

Thanks-

 

[DaveAtkins]

I used M = 7850 lb-ft (for the DL + WL case)
I used Fb = 1650 psi
I used S = 30.25 in^3

I also used
le = 19.2 ft (using 0.8 for K)
Fc = 1750 psi
Cf = 1.1 for Fc
Cf = 1.3 for Fb
Cr = 1.15
E = 1700000 psi

DaveAtkins