AS3600-2018 Core wall design for EQ - minimum reinforcement

[TFP_Byrd]

Aside from all the closed ligs, the new code's minimum reinforcement requirements for EQ also appear to have changed/increased:

We have a core box like this, with many internal walls:

The core is mainly in compression, with low tension forces, especially for the EQ cases. The wind cases do generate some tensions but they are still small.

I've been trying to find a way to justify that given the low tensions, there is no real "plastic hinge region", however, based on my reading online, plastic hinge zones can be defined regardless of whether the concrete element is in tension or compression.

I feel like I'm missing something though as the core will require N32-150 for the first 10 floors due to the core walls being thick. It doesn't make sense to have such heavy reinforcements given the tensions are so small.

Looking forward to the discussions!

 

[QSIIN]

If your wall isn't going into tension under the ductile assumption, then you have reduced your EQ load too much and assumed more ductility than the system actually has. Hence the code minimum of tension reinforcement to account for the over reduction.

You should check the tension requirements under non-ductile behaviour as a comparison, and compare other walls in the system to see how the dominant lateral load resisting element is behaving, if not the box. Based on what you've said above, the full earthquake load probably has significantly more tension than wind. Then you can determine if a non ductile or ductile design is most appropriate and efficient.

Also, depending on the thickness of the wall, N32-150ef can be very reasonable.

 

[HD-111]

Hi QSIIN, do you mean that if we do not receive any tension in the core wall with a μ of 2, we should instead assume non ductile behaviour with a μ of 1 even if the walls go into tension in that situation?

 

[Agent666]

Ductility works like this:-

You derive a design load based on your ductility assumption.

You have to match your design strength (and detailing) as closely as possible to actually achieve that ductility and form the selected mechanism.

If you provide twice as much strength as your analysis is reporting, then the ductility is more or less halved. If you provide 20% more then the design load is in fact 1/(1-0.2) higher in simplistic terms (obviously with the changes to Sp factors at lower ductilities and so forth the relationship is not linear)

Ductility uses up every bit of capacity you provide to form the ductile mechanism. The applied load is not capped at the design base shear at whatever ductility you selected. It will invariably be higher unless your strength matches perfectly.

I fail to believe you can have a ductile mu=2 building without a flexural hinge at the base of your core wall. If you have that you have tension at the capacity of the reinforcement (not low as assumed) on certain walls. Otherwise how are you dissipating the energy in your ductile mechanism unless the wall bars are yielding at some load corresponding the capacity provided??

As a New Zealand designer, there seems to be a general failing evident in a lot of the way this seems to be educated to engineers in Australia. Ductility should not be seen only as a means of lowering the load and then designing elements as if they were elastic. Once you ar einto ductility, you're accepting your members will yield at the capacity you provide. For a flexural wall/core where the axial loads are generally low, this means the reinforcement will be yielding to form a flexural hinge.

https://engineervsheep.com

 

[Retrograde]

If you are looking in Section 14.6 it means you are designing your walls as "limited ductile structural walls" or "moderately ductile structural walls". There is no way around the minimum vertical reo requirement.

If your tensions are so low, try designing the building/walls as "non-ductile" instead of "limited ductile". Then the minimum reo requirements of 14.6.7 will not apply.

 

[TFP_Byrd]

Thanks for all the responses.
What's actually driving the design at this stage is the period of the building.
Even when we run our model at u=1, the tensions are low, but the period of the building increases to much more than we typically find acceptable, which is approximately 0.1s for every 3m of height.
So we are hoping to maintain the u=2 but find a way around the higher minimum reinforcement requirements given the tensions are low. Is there a way we can justify that a plastic hinge isn't being developed and hence we can use the 0.0025 minimum?

 

[Retrograde]

The ductility factor should not have any affect on the period of the building. Can you explain what is happening here?

If you are using u=2 you cannot justify that a plastic hinge isn't forming.

 

[TFP_Byrd]

Okay, thanks for your response.
You're right - period is unchanged. I meant to say that the drifts are over the limit.

 

[Retrograde]

The drift should be roughly the same regardless of what ductility factor you use. If you refer to Section 6.7.2 of AS1170.4 you need to multiply your calculated deflections by u/Sp to get your design deflections. Sounds like your walls are not thick enough (or you need more walls).

 

[QSIIN]

HD-111, Section 11 tells you to do this.

It doesn't make any sense to choose "limited ductile" or "moderately ductile" if you're building is not behaving in a ductile manner, ie, not going into tension and going beyond the elastic limit.

Agent666, pretty much the majority of buildings in Australia are designed this way, so if we ever get an earthquake here, theres gonna be a lot of catastrophic failure.

TFP_Byrd
Your period can affect your EQ loads of you're using a modal analysis result in the static load calculation, but you can only reduce to 70% of the code equation. But if you're doing this, your model needs to include all elements to ensure you're not underestimating the stiffness, hence the 70% limit. If you're cracking your walls to reduce stiffness, but your walls aren't even going into tension, this is the wrong approach.

 

[rapt]

N32 @ 150 each face makes it a 750mm thick wall at f'c = 100MPa by my back calculation from the 14.6.7 equation. is that what you are dealing with?

If the walls are in low tension at mu = 2, they must have very significant tension at 2.7 times that load/sway with mu = 1.

If the section does crack, the 14.6.7 requirement is the minimum reinforcement required to generate more than 1 crack at the plastic hinge region and ensure ductile action. If it is not going to crack, you have to design as non-ductile anyway, so mu = 1.

 

[Agent666]

Agent666, pretty much the majority of buildings in Australia are designed this way, so if we ever get an earthquake here, theres gonna be a lot of catastrophic failure.

The only saving grace is perhaps your seismicity is very low and we aren't letting any of you buggers into NZ at the moment to work here, so we're safe for the time being.

We thought the low seismicity thing of CHCH, and half the city centre is still literally vacant lots 10 years later ..... in the words of Ozzy Man... destination f$%ked.


To OP, if your drifts are over the limit, you need a stiffer structure, it's that simple. Retrograde is right regarding scaling by mu, ultimate deflections irrespective of ductility are similar. If you are not scaling the analysis drifts by the ductility then there is a fundamental problem.

Chase your tail until it works, add more walls, increase stiffness, increase seismic loads, add more structure, increase stiffness, increase seismic loads. You'll get to a point where it works eventually. Then decide on ductility based on the strength you can achieve based on how much or little rienforcement you need to put in to satisfy all the constraints.

To put the ductility thing in perspective, Auckland has a Z of 0.13, which is our lowest, this is close to your highest Z (based on hazy memory of working on Australian jobs many moons ago). In Auckland almost every structure is designed for nominally ductile loads (mu=1.25). In 2011 CHCH event where a lot of structures were designed for ductility, all the concrete ones have been demolished. Concrete and significant earthquakes basically make your building a one time affair, achieves code intent of not collapsing and turning people into pancakes, but uneconomic to repair (not only structural cost, fitout and envelope damage played a big part). Basically our starting point in NZ is design as elastic as possible, then work towards ductility if warranted. It's not until you get to about a ductility of 2 that you get any benefit of lower loads on your foundations after consideration of overstrength. The extra design effort for a ductile structure and undertaking a capacity design often isn't worth it.

https://engineervsheep.com

 

[rapt]

Agent666,

I know of one who flew into Auckland in January, so you are not completely safe! He is currently being re-educated.

Yes, current seismicity in Australia is currently relatively low, but then it was in CHCH for a long time too. Historically, they are finding that Aus is far more active than everyone thinks, hence the changes to the current Earthquake Code and Concrete Code Earthquake rules. And we have had a couple of 7's in the last 30 years but they were in basically uninhabited areas so no problems. If one happened in a major population area, CHCH would look like a minor problem!

Now we just have to educate the designers, most of whom get very little education in this area and work under senior engineers who have very little experience in it and with builders who want to do it the way they always have. Good to see that some Aus engineers on this website seem to be learning!

 

[joeycheng]

Just wanted to add something in here. The comment regarding ductility no affecting period is not correct. Higher ductility assumes that a plastic hinge is forming in the building therefore causing a reduction in stiffness of the overall structure - this inherently means that the period of the building is increasing. This is accounted for in the increase of drift and lower forces when you adopt a higher ductility factor.

 

[Agent666]

Irrespective of the ductility you should be assessing the period under the same ultimate cracked stiffness, i.e. structure will attract load until it is cracked irrespective of what design ductility was assumed for the design. In NZ we are required to assess this stiffness at the point of first yield.

In a displacement design sense, if you're talking about the effective period then this depends on the ductility. But this isn't the value you're using to determine your seismic actions in a force based design.

https://engineervsheep.com