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HSS Connection
- Thread starter EngrRC
- Start date
Let's see if I understand why you do not weld around the perimeter. Do you do it to not transmit a moment? Do you want a shear connection?
"By feeling", I think that the welds would take linear stresses that could cause cracks in the ends under important loads. If you want a shear connection I would look for another option.
"By feeling", I think that the welds would take linear stresses that could cause cracks in the ends under important loads. If you want a shear connection I would look for another option.
- Thread starter
- #4
Hi DomDunk, I want it not to transmit major moment. When modeled so, it would give insanely huge major moments that I would not be able to design the connections. I want it to resist forces in 3 directions, as well as torsion. So with the detail attached, I could resist said forces, but it would also attract minor moment which is just fine.
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- #5
Lomarandil
Structural
I sure wouldn't call that a pinned connection with regard to major axis bending unless the end rotations were very very small.
I've looked in the past for ways to quantify weld ductility and make something like that work.. you can sometimes make it work with an analogy to a pre-qualified connection from AISC 9-14, but those welds tend to be really short and beefy (impractically so for light weight construction) -- not the full height like you've shown.
The one time I made it work... The contractor loved me, but man did those welders hate me....
----
The name is a long story -- just call me Lo.
I've looked in the past for ways to quantify weld ductility and make something like that work.. you can sometimes make it work with an analogy to a pre-qualified connection from AISC 9-14, but those welds tend to be really short and beefy (impractically so for light weight construction) -- not the full height like you've shown.
The one time I made it work... The contractor loved me, but man did those welders hate me....
----
The name is a long story -- just call me Lo.
Check this out,
I was surprised .... What do you think?
I was surprised .... What do you think?
- Thread starter
- #11
Thanks,
I can not calculate the stifness of base plates. But look at the results of the connection I sent you. It seems that effectively that type of connection is pinned ...
It would be good to check the rotation capacity of the connection with respect to what is required in your structural model with pinned connection?
I can not calculate the stifness of base plates. But look at the results of the connection I sent you. It seems that effectively that type of connection is pinned ...
It would be good to check the rotation capacity of the connection with respect to what is required in your structural model with pinned connection?
Lomarandil
Structural
That's interesting Daniel... I'm guessing the criteria are per Eurocode?
For this connection, fixity into concrete will reduce some flexibility, and the beam proportions will reduce some... But I'd guess not enough to change the result. I'll be curious if you are able to run it again and see the change.
----
The name is a long story -- just call me Lo.
For this connection, fixity into concrete will reduce some flexibility, and the beam proportions will reduce some... But I'd guess not enough to change the result. I'll be curious if you are able to run it again and see the change.
----
The name is a long story -- just call me Lo.
Yes,
it is the eurocode's criteria, with the AISC materials and parameters.
I thought I could not calculate the stiffness for base plate ... I was wrong.
Result: semi-rigid
it is the eurocode's criteria, with the AISC materials and parameters.
I thought I could not calculate the stiffness for base plate ... I was wrong.
Result: semi-rigid
EngrRC:
If you really want to transmit less moment to the concrete support structure, you will have to increase the stiffness of your RHS steel beam, so that it tends to induce less rotation at it end connections. And, you do not do that by trying to fiddle with the end connection welding details. In the extreme, substitute a 36WF for your light RHS. You will have essentially zero beam deflection and end rotation, so you just won’t be transmitting much potential fixed end moment, just gravity load shear and torsion which are fairly uniform You are either going to pull the top bolts out of the conc.; or the end pls. might be light enough (flexible enough) to bend enough btwn. the bolt line and the top (tips) of the side welds; or you are going to start to unzip the side welds at their top tips. The stresses in the last few inches of those welds will be very high when you add any moment component to the unit length average stresses from gravity load shear, torsion, and axial beam tension. The shear and torsion will cause fairly uniform stresses over the lengths of those welds, but the beam axial tension will cause some end pl. bending and prying on the welds and particularly the weld tips. Then you add the fixed end moment component potential and some end pl. bending and prying on the welds, again, particularly the weld tips, and you have a very high combined stress at the weld ends (tips). Then, add to that, that the weld ends with their starts and stops are the most critical (QC sensitive) areas on the whole weld, and that leads to a really nasty weld condition in your detail. As a means of thinking about and studying those kinds of details, look at how the various components tend to flex, rotate, deflect, etc., and wherever you get the max. interactions of these movements is also likely a very high stress area, and a likely failure initiation location.
If you really want to transmit less moment to the concrete support structure, you will have to increase the stiffness of your RHS steel beam, so that it tends to induce less rotation at it end connections. And, you do not do that by trying to fiddle with the end connection welding details. In the extreme, substitute a 36WF for your light RHS. You will have essentially zero beam deflection and end rotation, so you just won’t be transmitting much potential fixed end moment, just gravity load shear and torsion which are fairly uniform You are either going to pull the top bolts out of the conc.; or the end pls. might be light enough (flexible enough) to bend enough btwn. the bolt line and the top (tips) of the side welds; or you are going to start to unzip the side welds at their top tips. The stresses in the last few inches of those welds will be very high when you add any moment component to the unit length average stresses from gravity load shear, torsion, and axial beam tension. The shear and torsion will cause fairly uniform stresses over the lengths of those welds, but the beam axial tension will cause some end pl. bending and prying on the welds and particularly the weld tips. Then you add the fixed end moment component potential and some end pl. bending and prying on the welds, again, particularly the weld tips, and you have a very high combined stress at the weld ends (tips). Then, add to that, that the weld ends with their starts and stops are the most critical (QC sensitive) areas on the whole weld, and that leads to a really nasty weld condition in your detail. As a means of thinking about and studying those kinds of details, look at how the various components tend to flex, rotate, deflect, etc., and wherever you get the max. interactions of these movements is also likely a very high stress area, and a likely failure initiation location.
The stiffness evaluation is neat.
I have a concern for this specific situation with regards to the plastic redistribution of weld capacity assumption. I'm not convinced that the ductility exists in this style of weld connection. Research seems to show that single sided fillet welds on an HSS wall can't actually develop the full 90 degree from weld axis capacity that one would hope for and that failure happens closer to the theoretical parallel shear capacity of the weld, due to some combination of rotation, eccentricity and other effects. Given that it's not a pure yield failure, I would be wary assuming that this geometry of weld will redistribute forces plastically and not just unzip as it fails.
This might not be as big a deal with a connection with a top and bottom weld, since your capacity and stiffness are less impacted by the assumption the more you gather resistance towards the outside edges.
When I've tried to work this situation out before, and it's critical, I've done a couple of things. One was welding angles to the sides of the member and then welding the toes of the angle and some return to the connected element. It introduces some flexibility in the bending direction and still gives you some torsional restraint. You can also get creative with plates using various bolt slot directions. That's a bit trickier with HSS than it is with wide flanges, though.
I have a concern for this specific situation with regards to the plastic redistribution of weld capacity assumption. I'm not convinced that the ductility exists in this style of weld connection. Research seems to show that single sided fillet welds on an HSS wall can't actually develop the full 90 degree from weld axis capacity that one would hope for and that failure happens closer to the theoretical parallel shear capacity of the weld, due to some combination of rotation, eccentricity and other effects. Given that it's not a pure yield failure, I would be wary assuming that this geometry of weld will redistribute forces plastically and not just unzip as it fails.
This might not be as big a deal with a connection with a top and bottom weld, since your capacity and stiffness are less impacted by the assumption the more you gather resistance towards the outside edges.
When I've tried to work this situation out before, and it's critical, I've done a couple of things. One was welding angles to the sides of the member and then welding the toes of the angle and some return to the connected element. It introduces some flexibility in the bending direction and still gives you some torsional restraint. You can also get creative with plates using various bolt slot directions. That's a bit trickier with HSS than it is with wide flanges, though.
To add more to dhengr's point, a wide flange is fairly easy to design for this condition and not just throw beam stiffness at the problem. You can detail this as follows to restrain against torsion and not strong axis bending:
Connect the web for shear using a typical shear only welded or bolted connection with some flexibilitiy
Connect the flanges to the support using bolted cover plates with slots in the direction longitudinal to the beam.
Ensure that you clearly label the intent of the connection and direct that it not be slip critical, so a helpful erector doesn't try to tension the hell out of the bolts.
Of course, with a wide flange you don't have all the torsional strength and stiffness you're probably looking for, so this may not be a solution
Connect the web for shear using a typical shear only welded or bolted connection with some flexibilitiy
Connect the flanges to the support using bolted cover plates with slots in the direction longitudinal to the beam.
Ensure that you clearly label the intent of the connection and direct that it not be slip critical, so a helpful erector doesn't try to tension the hell out of the bolts.
Of course, with a wide flange you don't have all the torsional strength and stiffness you're probably looking for, so this may not be a solution
BridgeSmith
Structural
If you're aiming for a connection with high shear and axial capacity but limited moment transfer, you may want to try something with a elastomeric bearing pad under (behind?) a rigidly connected end plate. The closer you keep the bolts to the axis of rotation, the less moment you'll have.
HelloEngrRC said:
I hope the attached detail could be sufficient.
The top is the front view.
The bottom is the plan.
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