I was asked to check if the columns of a 30-story condominium building are properly designed to today's standards to carry the loads they are supposed to. I can use the ASCE 7-22 for the live load reduction etc, but I am not sure if I should just stick to a 40 psf load for Live load value that is the minimum we recommend. I never design for minimum even though I know that statistically that is already actually very conservative. Sure we have 100 psf for corridors, but even with that, in a plan that looks more like a labyrinth with corridors going around and units in really odd forms, I wonder if I could get a general idea of what is a good average value to consider. With design specifications for f'c of 3ksi, I am skeptical about how conservative I should be. Any suggestions?
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Recommendations for live load value of existing residential highriser 2
- Thread starter Iasonasx
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slickdeals
Structural
Corridors serving residential units need to only be designed to 40 psf. However, if the corridor is connecting an assembly area, that would be 100 psf.
I am not sure why you would be using 3 ksi concrete for a 30-story building. Are you sure it's 3 ksi? That seems a bigger problem than live load reduction.
What is your role on this project? Who is asking you to check this?
I am not sure why you would be using 3 ksi concrete for a 30-story building. Are you sure it's 3 ksi? That seems a bigger problem than live load reduction.
What is your role on this project? Who is asking you to check this?
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jerseyshore
Structural
Florida high rises and questionable designs, name a more iconic duo.
milkshakelake
Structural
I would design for live load of 40 psf. I personally design corridors to 100 psf, but that's above code requirements if it's only connecting to condos, and not necessary. I know about the 2 sigma rule for LRFD and how loads are made by the code committees; it's just a personal preference for me to overdesign corridors.
If the columns and foundations aren't working, use live load reduction for the columns. There is a footnote in Table 4-1 that live load reduction can't be used on "public rooms," though. I interpret that as recreational rooms, laundry rooms, and things like that.
I believe that the design should work if it was done correctly. If I recall correctly, the prevailing load combination used to be 1.3D+1.7L, or something like that. And 40 psf live load for residential has been used since the 1930's; it wouldn't be something higher.
If the columns and foundations aren't working, use live load reduction for the columns. There is a footnote in Table 4-1 that live load reduction can't be used on "public rooms," though. I interpret that as recreational rooms, laundry rooms, and things like that.
I believe that the design should work if it was done correctly. If I recall correctly, the prevailing load combination used to be 1.3D+1.7L, or something like that. And 40 psf live load for residential has been used since the 1930's; it wouldn't be something higher.
40 psf for a corridor isn't code anywhere I've looked. Further, egress path is 100 psf. Can anybody clarify that?
I would honestly be more concerned with the lateral, but as you dig through existing drawings, depending on the time frame it was constructed, certain "common flaws" should be given due consideration, columns can be fairly quickly checked for overcrowding, but how do you address the problem?
I'm not sure what you're getting checking it per the current code, particularly the current material code.
But the classics, if this is Florida, circa 1980 (Champlain Towers, Surfside, Harbour Cay, Cocoa Beach) had slabs that were thinner than that required for waiving deflection checks, slab-column joints that didn't typically have enough reinforcement, punching shear neglected or not calculated, and reinforcement in the columns beyond the traditional 4% limit that allows "normal" splices without exceeding the 8% gross area limitations. I don't have any plans for Harbour Cay, but Champlain towers had shear walls that didn't look like there were enough. The columns farther down one would prefer to have the same cross section and they change the reinforcement and concrete strength. Ration of slab strength versus column strength can also crop up as a violation, particularly at the lower levels where the column strength is higher but "nobody" increases the slab strength to meet the requirements. Not sure how to fix that, either.
Back in the day (Sap90) at least some designers would run all the lateral frames side by side and connect them with pinned connectors to force all the frames to deflect the same amount, so that abstracts the diaphragm force distribution (the analysis can't provide it), and while it's a rigid diaphragm, there are zero differential deflections between the frames, beyond whatever axial shortening happens in the slabs, and it doesn't capture any torsion due to an irregular structure or an irregular shear wall or moment frame arrangement. Older wind codes didn't really do any eccentricity, or didn't always.
Keep in mind the spreadsheet tools were limited at the time and a lot of things were handled in a more simplified fashion, so rather than going perhaps "pure calculated tributary area" they may have done a certain percent of live load reduction per floor and beyond say, the top three floors, used the maximum reduction.
I wouldn't put faith in anything being done correctly in the 1980s. I'll just say Miami Vice and call it a day.
Regards,
Brian
I would honestly be more concerned with the lateral, but as you dig through existing drawings, depending on the time frame it was constructed, certain "common flaws" should be given due consideration, columns can be fairly quickly checked for overcrowding, but how do you address the problem?
I'm not sure what you're getting checking it per the current code, particularly the current material code.
But the classics, if this is Florida, circa 1980 (Champlain Towers, Surfside, Harbour Cay, Cocoa Beach) had slabs that were thinner than that required for waiving deflection checks, slab-column joints that didn't typically have enough reinforcement, punching shear neglected or not calculated, and reinforcement in the columns beyond the traditional 4% limit that allows "normal" splices without exceeding the 8% gross area limitations. I don't have any plans for Harbour Cay, but Champlain towers had shear walls that didn't look like there were enough. The columns farther down one would prefer to have the same cross section and they change the reinforcement and concrete strength. Ration of slab strength versus column strength can also crop up as a violation, particularly at the lower levels where the column strength is higher but "nobody" increases the slab strength to meet the requirements. Not sure how to fix that, either.
Back in the day (Sap90) at least some designers would run all the lateral frames side by side and connect them with pinned connectors to force all the frames to deflect the same amount, so that abstracts the diaphragm force distribution (the analysis can't provide it), and while it's a rigid diaphragm, there are zero differential deflections between the frames, beyond whatever axial shortening happens in the slabs, and it doesn't capture any torsion due to an irregular structure or an irregular shear wall or moment frame arrangement. Older wind codes didn't really do any eccentricity, or didn't always.
Keep in mind the spreadsheet tools were limited at the time and a lot of things were handled in a more simplified fashion, so rather than going perhaps "pure calculated tributary area" they may have done a certain percent of live load reduction per floor and beyond say, the top three floors, used the maximum reduction.
I wouldn't put faith in anything being done correctly in the 1980s. I'll just say Miami Vice and call it a day.
Regards,
Brian
@Iasonasx ,
If the building is 30-story building with concrete strength of 3ksi, my concern would be the low strength of concrete rather than looking for LL reduction factors.
My opinion..
Use it up, wear it out;
Make it do, or do without.
NEW ENGLAND MAXIM
If the building is 30-story building with concrete strength of 3ksi, my concern would be the low strength of concrete rather than looking for LL reduction factors.
My opinion..
Use it up, wear it out;
Make it do, or do without.
NEW ENGLAND MAXIM
- Thread starter
- #8
@HTURKAK,
Of course that is my concern too, just one of them ...
I just don't see that one concern is all I should have in such a case. There are many issues I have to address. If in the end it is designed to work with the 3ksi, well good for all of us. I don't mind if they have larger columns that compensate for the lesser strength. I will have to address issues of maintenance and condition of the structure too. So I started by asking what would be a more standard value colleagues use for such cases for average live load.
Of course that is my concern too, just one of them ...
I just don't see that one concern is all I should have in such a case. There are many issues I have to address. If in the end it is designed to work with the 3ksi, well good for all of us. I don't mind if they have larger columns that compensate for the lesser strength. I will have to address issues of maintenance and condition of the structure too. So I started by asking what would be a more standard value colleagues use for such cases for average live load.
- Thread starter
- #9
@ lexpatrie,
I agree with all you say except of the "I'll just say Miami Vice and call it a day". There are people living in these structures and they paid top dollar to buy them. I have to do what I have to do and give them a substantiated set of comments about their building.
I agree with all you say except of the "I'll just say Miami Vice and call it a day". There are people living in these structures and they paid top dollar to buy them. I have to do what I have to do and give them a substantiated set of comments about their building.
milkshakelake
Structural
@lexpatrie From ASCE 7-10 (hopefully not too outdated):
milkshakelake
Structural
Also, I agree with lexpatrie's comment about structural. Some older seismic codes had pretty wild R values that were way too high.
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jerseyshore
Structural
This is quite the slippery slope these days of existing structural analysis. I understand why these larger buildings are being re-checked, especially in Florida, but where does it end? NJ just passed a bill for required condo analysis too. The liability seems incredibly high on this work.
And realistically what percentage of buildings 30+ years old are going to have elements that aren't up to today's code standards? I'm sure quite a bit. If you tell them these columns aren't up to code are they going to tear the building down? Evacuate? Reinforce the whole thing?
And realistically what percentage of buildings 30+ years old are going to have elements that aren't up to today's code standards? I'm sure quite a bit. If you tell them these columns aren't up to code are they going to tear the building down? Evacuate? Reinforce the whole thing?
- Thread starter
- #14
@jerseyshore
You're right. It is a slippery slope. I intend to just do my job and if I tell them that the numbers yield that this structure is inadequate according to today's standards, it is up to them to take action according to their decision making standards. I fear that I will find a few columns to be inadequate for what they need to support. I just intend to do my job ethically and provide the appropriate report.
You're right. It is a slippery slope. I intend to just do my job and if I tell them that the numbers yield that this structure is inadequate according to today's standards, it is up to them to take action according to their decision making standards. I fear that I will find a few columns to be inadequate for what they need to support. I just intend to do my job ethically and provide the appropriate report.
structSU10
Structural
Take a look at ACI 562. They deem something unsafe at a demand/capacity ratio of 1.5. I have only skimmed it for general information but it may have something useful to tie into your assessment.
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I didn't mean Miami Vice as some flippant comment. I meant it as a sort of oblique comment regarding people's judgement being affected, and the quality of design in Florida in the 1980s that I've seen documented. I meant don't take anything as a given that it's properly designed, particularly if you have actual construction drawings. Missing rebar, terrible inspections, etc. To say nothing of poor design and people not bothering to check their work.
How these hypothetically defective structures get ammended is on us, as the living engineers. Life safety and welfare. If you can't figure out how to reinforce it, I think that's "fine" so long as you can produce a coherent report that will allow another engineer to undertake the work (after confirming you're not a wingnut).
Where does the line get drawn between this is a) defective b) dangerous, c) hazardous d) wrong but fine and e) evacuate now, get drawn? That I don't know. It depends on each individual engineer and what their decision making process is. I'll point at Davenport, Iowa, however. Clearly at least some engineers can't see a hazard, or won't. New York City is another example. At least one South Florida condo was evacuated and extensive repairs to the loadbearing columns was performed. I won't name it, because the other one the guy looked at got minimal repairs and collapsed.
Going back to the Miami Vice comment, corridors are exit ways. Horizontal exit ways, transfer corridors, whatever you want to call them.
Corridors, the way a lot of these places are designed, the first floor is pretty public, and it's already 100 psf as a first floor corridor. The upper floors I see a lot of folks wanting to use quite a bit less, it doesn't do much for the column design on a multi-story building one way or the other, and in a concrete building it's pretty impractical to reduce the depth if the live load is 40 versus 100, only for the corridor, and I'll stand by the 100 psf. I've seen projects kicked back by more rigorous building departments for using 40 psf in a "private corridor" in a hotel when it's actually for accessing the rooms from the public elevator and the egress stairs. A private corridor is inside a private room.
That doesn't mean that some yahoo engineer didn't design something in 1980 for 40 psf live load and nobody bothered to review it, or for that matter, file the drawings. It isn't all that realistic for, say, a hotel corridor to experience 100 psf, but that's what the code (life safety) calls for, as I've seen it interpreted, so if you're evaluating a design, I think you have to expect all of it to be done wrong and then work to disprove that assumption via generally accepted principles of mechanics and structural analysis and the code in place at the time of construction/design.
It's fairly likely a 33 story concrete building that was designed incorrectly (quotation marks deliberately not provided) for 40 psf exit ways wouldn't be gravely affected by this error (provided you even agree with me on 100 psf, which I rather doubt).
People design stuff for forty years (or twenty) and somebody comes along and asks a question and there's no possible way the other person has been doing it wrong for twenty or forty years, so it gets discarded. Ego protection. Or you get the "ultimately it won't matter" hand waving argument and no calculations. It's fine to disagree, but the hand-wave is what I dislike, if it doesn't matter, put down the calculations SHOWING that. It reminds me of the old "design for 10 psf net uplift" notes I'd see on steel joist framing plans. There's never a calculation showing that 10 psf applies (and it can fairly often be inadequate, the structure just happens to not see the design wind speed, SO FAR).
Some of these checks are exceedingly easy to perform, like checking percent reinforcement in a concrete column, the floor strength/column strength limit, the minimum depth to ignore deflections in a two-way slab, this is all so trivially easy to do and you look at Harbour Cay and Champlain Towers South and all three of these items are a fail.
When stuff this basic cannot be depended on being done correctly, the entire structure must be viewed with extreme skepticism. Admittedly, Harbour Cay was designed by two people with, I believe, literally no training or education in the field of concrete design, but it isn't like these checks are somehow not in the code if you read it from cover to cover. It's all there in the printed sections, written with visible ink. No secret decoder ring needed. You would think somebody with zero design experience would be MORE literal with the code, not less. And yet, no.
How these hypothetically defective structures get ammended is on us, as the living engineers. Life safety and welfare. If you can't figure out how to reinforce it, I think that's "fine" so long as you can produce a coherent report that will allow another engineer to undertake the work (after confirming you're not a wingnut).
Where does the line get drawn between this is a) defective b) dangerous, c) hazardous d) wrong but fine and e) evacuate now, get drawn? That I don't know. It depends on each individual engineer and what their decision making process is. I'll point at Davenport, Iowa, however. Clearly at least some engineers can't see a hazard, or won't. New York City is another example. At least one South Florida condo was evacuated and extensive repairs to the loadbearing columns was performed. I won't name it, because the other one the guy looked at got minimal repairs and collapsed.
Going back to the Miami Vice comment, corridors are exit ways. Horizontal exit ways, transfer corridors, whatever you want to call them.
Corridors, the way a lot of these places are designed, the first floor is pretty public, and it's already 100 psf as a first floor corridor. The upper floors I see a lot of folks wanting to use quite a bit less, it doesn't do much for the column design on a multi-story building one way or the other, and in a concrete building it's pretty impractical to reduce the depth if the live load is 40 versus 100, only for the corridor, and I'll stand by the 100 psf. I've seen projects kicked back by more rigorous building departments for using 40 psf in a "private corridor" in a hotel when it's actually for accessing the rooms from the public elevator and the egress stairs. A private corridor is inside a private room.
That doesn't mean that some yahoo engineer didn't design something in 1980 for 40 psf live load and nobody bothered to review it, or for that matter, file the drawings. It isn't all that realistic for, say, a hotel corridor to experience 100 psf, but that's what the code (life safety) calls for, as I've seen it interpreted, so if you're evaluating a design, I think you have to expect all of it to be done wrong and then work to disprove that assumption via generally accepted principles of mechanics and structural analysis and the code in place at the time of construction/design.
It's fairly likely a 33 story concrete building that was designed incorrectly (quotation marks deliberately not provided) for 40 psf exit ways wouldn't be gravely affected by this error (provided you even agree with me on 100 psf, which I rather doubt).
People design stuff for forty years (or twenty) and somebody comes along and asks a question and there's no possible way the other person has been doing it wrong for twenty or forty years, so it gets discarded. Ego protection. Or you get the "ultimately it won't matter" hand waving argument and no calculations. It's fine to disagree, but the hand-wave is what I dislike, if it doesn't matter, put down the calculations SHOWING that. It reminds me of the old "design for 10 psf net uplift" notes I'd see on steel joist framing plans. There's never a calculation showing that 10 psf applies (and it can fairly often be inadequate, the structure just happens to not see the design wind speed, SO FAR).
Some of these checks are exceedingly easy to perform, like checking percent reinforcement in a concrete column, the floor strength/column strength limit, the minimum depth to ignore deflections in a two-way slab, this is all so trivially easy to do and you look at Harbour Cay and Champlain Towers South and all three of these items are a fail.
When stuff this basic cannot be depended on being done correctly, the entire structure must be viewed with extreme skepticism. Admittedly, Harbour Cay was designed by two people with, I believe, literally no training or education in the field of concrete design, but it isn't like these checks are somehow not in the code if you read it from cover to cover. It's all there in the printed sections, written with visible ink. No secret decoder ring needed. You would think somebody with zero design experience would be MORE literal with the code, not less. And yet, no.
I've always considered the minimum code values as being more than adequate, even with the live load reduction factors. There are a few bits of literature stipulating the 'actual' live load values and they are a fraction of the reduced live load. I've never felt uncomfortable in using the code stipulated loadings.
-----*****-----
So strange to see the singularity approaching while the entire planet is rapidly turning into a hellscape. -John Coates
-Dik
Granted.
Perhaps you mean the 1950s live loads in buildings from the NBS? Live Loads on Floors in Buildings, 1952, Dunham, Brekke, Thompson?
Trans. ASCE Design Live Loads in Buildings, Dunham, Jan 1947
Hypothetically at least one person manually did live load reduction as suggested by Ziemian and McGuire, A Method for Incorporating Live Load Reduction Provisions in Frame Analysis, Engineering Journal, first quarter, 1992. I don't think this was ever blatantly incorporated into any software around that time, not that I'd really be in a position to know definitively.
Analysis of Live Loads in Office Buildings, Ellingwood, Culver, J. Struct. Div, Aug. 1997 (this was difficult to find, I thought it was Ellifrit)
Apartments I suppose haven't changed that much, waterbeds aren't much a thing anymore, and the TVs are getting bigger but lighter. The "paperless" office is probably another thing that's driving actual live loads down. You can get insight into "actual" office live loads in a lot of the vibration design guides as you get incorrect damping if you use the full live load.
Regards,
Brian
Perhaps you mean the 1950s live loads in buildings from the NBS? Live Loads on Floors in Buildings, 1952, Dunham, Brekke, Thompson?
Trans. ASCE Design Live Loads in Buildings, Dunham, Jan 1947
Hypothetically at least one person manually did live load reduction as suggested by Ziemian and McGuire, A Method for Incorporating Live Load Reduction Provisions in Frame Analysis, Engineering Journal, first quarter, 1992. I don't think this was ever blatantly incorporated into any software around that time, not that I'd really be in a position to know definitively.
Analysis of Live Loads in Office Buildings, Ellingwood, Culver, J. Struct. Div, Aug. 1997 (this was difficult to find, I thought it was Ellifrit)
Apartments I suppose haven't changed that much, waterbeds aren't much a thing anymore, and the TVs are getting bigger but lighter. The "paperless" office is probably another thing that's driving actual live loads down. You can get insight into "actual" office live loads in a lot of the vibration design guides as you get incorrect damping if you use the full live load.
Regards,
Brian
milkshakelake
Structural
@lexpatrie Thanks for that snippet. It's interesting to me because my local code calls it "Stairs and exits," not "Stairs and exit ways." It changes the meaning quite a bit. I just sent a request for clarification to the building department, and I'll update here in the small chance that I get a response. I feel like if they wanted 100 psf live load for the corridors that serve as exit ways, they would've said so in the corridors section. Because almost every corridor in this universe is an exit way, besides some fancy places that have internal corridors.
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