Shear Wall Hold Down Distance Apart

[CurlyQue]

When designing a wood shear wall with gypsum board or wood diaphragms, is there a maximum distance the hold-downs can be from each other? If a shear wall is 20' long versus 100' long, the uplift is more for the 20' than the 100' so longer shear walls give you better leverage to prevent overturning. However, 100' long walls seem more prone to buckling than 20'.

Thans

 

[CurlyQue]

AELLC per 'did some research' ...

I'm aware of the ATS system that compensates for wood shrinkage; the SE is not specifying this, just a standard all thread. He objected to the contractor even using straps instead because he said the straps would pop when the wood shrinks. So, he's aware of wood shrinking but doesn't accommodate it in the design of the all threads. I did put a note on the drawings I prepared for him to tighten the nuts before putting on gypsum board. That's the best I could do.

I'm saying the multi-family building is long, 150'x72' with exterior jogs every 12-20' and jogs at the midpoints interior to offset differences in units along a common-wall spine.

I see jogged shear walls as just being some walls oriented for wind in one direction and some for wind in the other. Offset shear walls aren't necessarily a problem, you just can't get as much length as ones that don't so the uplift is going to be higher. Some are only 8' long.

Thanks for your input.

 

[AELLC]

Well in that case the shear walls should have been in the range of 8' to 20' long.

Since the foundation is already in, why wasn't any red flag raised earlier?

 

[CurlyQue]

Cadair,

Thanks for your post.

I agree with a lot of what you're saying but ...

To your fist point" If they don't act as a unit then how can the overturning moment be resisted by two hold downs, one at each end of a 'long' wall. You wouldn't be able to use the lever arm of the entire length so your point supports the engineer's position to panelize You can sheathe shear walls in different ways: vertical, horizontal, or running bond. The panels are also nailed to common studs, sometimes two where nail spacing requires. As such, the wall does act as a unit although this specific engineer sees them as individual panels too.

To the second point: From a structural viewpoint, the wall is being loaded along it's length but you can also view it as a column laterally loaded from floor to floor, having a width of floor to floor, but having a depth of the thickness of the wall. Viewing it this way just means you're turning the problem on it's head (really on it's side) and viewing like it's a column to address the buckling issue. As far as I know, this is acceptable in structures. For example, a grade beam or footing when turned upside down structurally is like a uniformly loaded beam. That's not saying I'm going to turn it upside down when it's built, of course not, but you can analyze it that way. So too with this problem as I'm seeing it. Forces are forces and you can view them as if they're floating in space. Gravity is another force.

To your third point: You're point is will taken. It would seem as if the down force of the adjacent panel does counter the up force of the other. If so, then the two negate so the resultant is zero. How then is the overturning moment resolved; at the end points of a long wall as is traditionally viewed? The endmost panels have no negating force so, according to your viewpoint, the first panel loaded would bear the entire overturning moment and the uplift would be that force divided by the width of the panel which is what, a 4' panel? To follow your load trail then, the the force is sent from panel to panel and the downward force is taken up at the end panel, whatever it's 'width'.

To your last point: I agree. The hold-down forces are into the 'page', the foundation, the ground. I am talking about looking at a shear wall as if it is a column on it's side with the lateral force acting as if it's loading the top of a column on it's side as I said in point two.

Thank you for your thoughts.

 

[CurlyQue]

AELLC per 'well in that case':

I don't follow how you get to a range. Some are 24' long, some are 48' long, some are 8' long depending on the wall layout by the architect. There was no real thought about structure from the architect's position so the shear walls got shoe-horned.

Actually, the owner 'purchased' a design from an architectural firm in Texas to build in Indiana and the design-builder bid the project without structural drains. They ball-parked a figure then came to us (me) and said we need a structural design ASAP. We contacted an engineer who designed what we (I) documented. The design-builder then screamed at the size of the foundations saying, after 2 hours of review, that they were 'too-beefy'. They then screamed when the framing drawings and the infamous shear wall strategy came to light saying it's costing $160K more than their budget, whatever that is.

Isn't this how design-build works in your neighborhood?

The red flags weren't thrown because the foundation design and construction somewhat followed the framing design.

I think they call this SNAFU in army speak. The design-builder was expecting gypsum board shear walls but did not make this expectation known in advance; they assumed all shear wall are done this way. The engineer we hired apparently doesn't do those nor does he do 'long' walls. So, this project has been sort of a perfect storm and I'm in my dingy between calamitous waves.

So now you know the rest of the story.

 

[AELLC]

OK I see. I never was involved in design-build for wood MFH.

Then the shearwalls should have been 8', 24', or 48' long maximum, maybe shorter.

Is it understood that a jogged walls cannot be counted as one shearwall? Every shearwall needs to be one straight panel, with or without a window opening. One strap or hold down at each end. The straps dont buckle due to shrinkage because they "grab" onto the end of studs only the distance required for nails to transfer the load, say 12" to 24", and the strap passes over the floor truss depth with minimal nailing.

If it dodn't cost much to temporarily stop the construction, I would guess it a possibility to re-engineer this whole mess with epoxied-in hold downs to the foundations, but that may snag because of high forces at each hold down makes your SE nervous about the regular continuous wall footing wasn't designed for those forces.

 

[Cadair]

CurlyQue,

They don't act as a unit. The overturning moment can be resisted by two holdowns because the middle panels cancel each other out. It doesn't matter if the panels are vertical, horizontal, running bond, whatever. They are units. They will be stronger if they share a common stud as that is how the compression in one resists the tension in the other.

Second point. Okay, I think I'm seeing what you are talking about in the right orientation. Now how does a longer (as in longer wall) mean it is more likely to buckle? I'm assuming stress is constant with varying lengths of wall. How is I varying? Which plane is I in?

Third point. Basically, yes. The holdown is to restrain the first panel. The first panel restrains the next and so on. An easy, conservative way to size holdowns is to take the capacity of the shearwall in plf and multiply it by the height of the wall. The holdown has to be that strong. Notice that it has nothing to do with the length.

Basically, the point is that wood shearwalls don't act as units but rather as individual panels. You don't have to believe me though. Ask the people who write the standard, the American Wood Council.

http://www.awc.org

 

[CurlyQue]

AELLC per "OK I see." (and to JAE)

The shear walls are as you have stated, some short, some long.

It is my understanding that jogged walls cannot be counted as one shear wall as you stated. Where the wall jogs, that jog takes load from the 90 degree direction whereas the other walls take load from the 0 degree direction.

The straps don't buckle; they can't because they're anly used to resist tension, the uplift force. When the posts (wherever they are) are in compression then the straps can buckle somewhat due only to the fact that the compression member is compressed. The nails, in that case, simply keep the straps from buckling. The main point is that the straps are for tension only as I know you know.

Re-engineering did present itself. The question became: "Okay, if you only want gypsum, then we'll have to get another engineer." But the design-builder 'bought' into having OSB because they really didn't have the time to stop construction and wait for re-engineering. That delay would have killed their schedule. The owner needs the building done for rent by a certain date ... as usual.

It's all a case of inverse NASCAR: going too fast and slowing down or, how they put it: "You've got to slow down to go fast. That seems counter-intuitive but it's true even for design-building; things take time and pushing to fast creates all sorts of needless problems. But that seems to be the design-build industry it seems or, at least, that's been my woeful experience.

It's what I call 'The Conundrum' or, rather, 'The Conumbrain." Scream and yell but go faster and cheaper, ASAP!

My main reason for the post is to try to determine whether the structural engineer we used for this project has a point. Are 4', panelized, shear walls stronger than long, 'flimsy' ones. If so, why? I believe there is some merit in his approach but I don't agree with his math (divide total overturning moment by 12 panels and then by 4' per panel) because the resulting uplift is the same as if I simply divided by 48' and put two hold downs. If he were to sum his moments, instead, and do it the way I was taught, then his uplift forces would be on the order of 6.5x less which would be a good reason. But, if one looks at the panels as being 4' long panels with 12x less force then that's like saying I'm loading up a 4' 'column' with a 'downward' force instead of a 48' 'column' with more force. From a column perspective that means my L/r is less and the 'column', the wall, is much more 'stout.' From an axially loaded beam (a column on its side) that means the web is stiffer or has lateral web-stiffeners, the perpendicular walls as was discussed by woodman88.

I wanted to see how others view this engineer's approach. And, I'm trying to figure out if there is a mathematical approach that verifies this structural approach.

With due respect for everyone's answers and they are all valuable, so far, JAE's My/I and P/A is the closest I've come to what I consider a structural approach. I haven't made time to analyze it more closely but it's in line with what I was taught too many years ago.

Thanks again.

 

[CurlyQue]

Cadair,

Your points are well taken. I like that you have supported your position again and referenced another source. I will look at AWC ... again.

Given your position, how is a shear wall with hold-downs at 4' stronger than one with hold-downs at 28'. I'd say 100' but JAE would grimace.

I (not me) is in the horizontal plane just as you see 'I' <-- here. That is, look at the wall as being the web of an I beam. The beam is 10' long in the z axis projecting toward you (the height from floor to floor) the depth of the beam (the length of the shear wall) is say 40'. The flanges of the 'beam' are the loaded walls, that is, the catchment, the amount of wall receiving wind force. A building viewed this way, in plan, is nothing more than a bunch of I-beams side by side all 10' or 20' long or however high your building is. The lateral force applied to the side of the building is the same as the one applied to the 'top' of these beams. The I value is simply the area of all those walls (flanges and web) times their distances squared (important) from their centroids to a base point, the bottom of the I; how you figure all I values of any cross section square, I-shaped, C-shaped, any shape. Woodman88 says My/I which is the same as saying Fv=Mc/I, the total overturning Moment times its distance (c or y in My/I)from the centroid of the I value divided by the I value. You then can calculate what the actual shear force at any point in the I value which, in this case, is actually shaped like an I beam. Likewise, F=P/A which is the lateral load / the area of all the walls in the 'I-beam' in plan.

What I'm focusing on is how the wall is like the web of this 'I-beam'. It is laterally braced at it's ends (the floors)so it is restrained in that respect like a beam cast into a concrete wall. Apply the forces and you will see that the 'web', the shear-wall, will want to buckle if enough force is applied. This can be corrected by adding thickness to the web or providing 'web-stiffeners', the walls in the room perpendicular to the shear wall.

You can see all of these mechanisms by playing with a shoe-box. It is a good model except that your looking at two C shapes rather than one I shape. Take the lid off (un-restrain the flange), apply a force to one end and you can see how the 'web', the shear-wall, wants to buckle. What I'm focusing on is the web buckling with the lid still on. How it will want to bow. But that's all assuming the wall, the side of the box, wants to act as a unit which you and AWC say it does not because the box side is made of individual units and they 'tranfer' their loads adjacent to each other like soldiers shoulder to shoulder or that thing that has 6 or seven balls suspended and when you lift one and release it, it sends its force through to the unrestrained one at the other end and causes that ball to bounce upward.

This is what I've been taught.

The structural engineer we hired for this project sees things similarly but I haven't quite I haven't quite yet figured out what all is involved in analyzing a shear wall this way.

But, if what you and AWC says is correct then 'my' engineer is missing the point.

Thank you again.

 

[CurlyQue]

Cadair,

Looking at the building as a bunch of I-beams is somewhat misleading because the wind force is transferred through the stud of the side walls up and down to the horizontal floor plates, the plywood deck, and then to the top and bottom of the shear wall. The load is not transferred laterally along the wall to the end of the shear wall. I only say the load wall is like a flange of an I beam because it represents that load area; the loads get to the shear wall through the floors, just to be clear that you know I know how the loads get there.

If you look at the 'I' shape as being vertical same as a wall section, the concept is similar because the lateral force would the be applied in the z direction (toward or away from you) but the buckling would be the studs and the panel deflecting in the x direction (left or right).

Why I'm looking at it the other way is that the force is lateral but I'm trying to determine how that affects the length of the wall as if it were a web buckling.

The difference is that I can see the force if the I is in plan view whereas I have to imagine the force coming toward or away from me if the I section is in section view.

The point is that in plan view, the structural engineer is placing dots that represent hold down locations at 4' in the I section in plan and somehow this configuration is stronger than dots only at the intersection of the I.

What I see is that the dots are closer and that they are resisting lateral load like post-tensioning a concrete beam is stronger than a beam that is not post-tensioned. Put another way, the lateral load is divided up among the dots, the hold down locations, such that the force is applied to a 4' 'length' of web versus a 40' length of web; that is, the L/r, the length of the 4' wall has less than for a 40' wall. It is like a column (the web) is only 4' high versus 40' high so the buckling would be less.

My head is spinning from all this rotating so I'm sure others are quite thoroughly confused.

I'm just trying to understand why a structural engineer considers hold-downs at 4' on center is somehow stronger than one with hold-downs at 40'. That is all.

Thanks

 

[woodman88]

As for having holdowns spaced 4' oc. He may be considering the shearwall to be acting like a Perforated Shear Wall per the ANSI / AF&PA SDPWS-2008. See section 4.3.6.4.2.1 of it. You can get a free copy of it at

http://www.awc.org/Standards/SDPWS.html.

Garth Dreger PE - AZ Phoenix area
As EOR's we should take the responsibility to design our structures to support the components we allow in our design per that industry standards.

 

[AELLC]

Garth -

Where does it say in the Sect. 4.3.6.4.2.1 that those are hold downs at 4' oc? I interpreted it as anchor bolts at 4' oc and hold downs only at the ends of the entire shear wall, not individual segments (I have been using perforated technology since the APA TR-157 report first came out)

 

[woodman88]

AELLC - What it states is that the length of the full height sections of shear walls must be tied down for a plf uplift force. Do it 1' oc, 2' oc, 4' oc or whatever oc but do it.

Garth Dreger PE - AZ Phoenix area
As EOR's we should take the responsibility to design our structures to support the components we allow in our design per that industry standards.

 

[hello12345hello]

i have 3 story wood shear wall with 12'4" total length and it has 6ft opening in center. So fully sheathed wall segment is 3'-2" and 3'-2" each side of opening. total ht is 30ft. Can i still use perforated shear wall approach.

 

[Cadair]

Regarding, "Given your position, how is a shear wall with hold-downs at 4' stronger than one with hold-downs at 28'. I'd say 100' but JAE would grimace."

It isn't. All fully restrained shearwalls (of the same material and nailing) are equally strong on a plf basis. Longer walls just have more length, so they can resist more force.

Regarding the I-beam idea. Thanks for the explanation so that I could see clearly. I agree with the other posters that say if the wall is long enough, the diaphragm will brace it via the studs.

Woodman may have a point if it is a perforated shearwall, then additional holdowns will add strength because perforated shearwalls are not technically always fully restrained. 4' o.c. is a little overboard, but if he can't figure out which is or isn't fully restrained...

It still has nothing to do with buckling though.

 

[CurlyQue]

Cadair,

My hat's off to you.

I took your advice and called AWC. I talked with Loren Ross who was very enlightening. Wood shear walls are definitely not monolithic. The forces are in the 'chords' and the panels just keep the rectangles from becoming parallelograms.

I'm done pursuing bending in long walls. A wood shear wall will fail before that ever happens. If a hold-down fails, then the building basically crumples before it ever would overturn.

I asked Loren if this concept is in the AWC standard and he said it's not. Loren said he's working on a document which describes what he discussed.

Do you know of any document that describes how wood shear walls work using chord forces?

Thank you so much.

And thanks to everyone else too.

 

[oldrunner]

I haven't read all of the posts above but do have one question. Is each 4' panel section(either sheathed with sheet rock or OSB) stacked next to each other, but the sheathing does not overlap from panel to panel? What I'm trying to determine is whether you have a line of individual panel sections only connected together by some sort of top plate and either the sill plates at the 1st floor and then either single or double top plates at each floor, but no connection between the panels at the mating studs.

 

[jdgengineer]

Old runner I'm not sure I understand your question. Typically Shearwall construction is "blocked" so you will have nails around the perimeter of each panel with your edge nailing requirement. Assuming you have an 8' top plate each panel will be nailed to the top plate sole plate and adjacent studs. Where the two panels meet they will be nailed to a common stud to create continuity. If the nailing requirements are low this stud can be a 2x if they are more tightly nailed it's often a 3x.

 

[CurlyQue]

oldrunner and jdgengineer,

Thank you both for reading through all of the posts. I apologize for the length. My main intent was to see how a 4' panel system is stronger than a 'continuous' shear wall; why the first won't 'crumple like a piece of paper' and the second, supposedly, will.

Each 4' panel is sheathed with OSB. Because each is a panel, there are 2 studs so the panels are fully engaged to studs. The studs are not, however, necessarily nailed together; the forces aren't negated by each panel. Each panel works separately. Each panel has a hold down anchor bolt with a 3" square washer and straps that tie the top of one panel through the floor system above to the next panel above. Each 'full panel' is essentially 4'x30' which seems to conflict with h:w ratio.

Each panel is constructed in field and the studs are built as a typical wall but with double studs at each panel point. Structurally, though, the panels are not tied together with 12d or 16d at 4" o.c. or so.

The engineer insists that all panel points have to be double 2x. I think that's required for high load walls per code table but not for every wall; the code is somewhat ambiguous saying that plywood has to be nailed to a full 2x which to some means 2 edges on 1 stud.

We had another engineer look at the whole design because we were concerned that there might not be adequate hold downs at the ends. The panels were analysed independently. The result was that the design works ... a whole lot more than is required. The design uses 33% of its capacity in wind loading and 66% of its capacity in earthquake loading which in Indy area is very nominal. The overdesign is really a result of the engineer insisting on carrying the OSB all the way up through a 3 story building. While I'm not a big fan of gypsum board sheathing for shear walls and other engineers I've talked to have a similar opinion, gypsum board does have some value, approximately 20-25% of OSB.

The current debate is that the contractor or design-builder is over budget on these walls. They say they do all of their buildings with gypsum board but I maintain each building is unique. a 50' wide unit (this building) is not the same as a 25' wide one; sometimes gypsum board isn't enough. We talked about the design at the beginning of the year and gave them full opportunity to go with another engineer but they decided to go forward only to come back 3 months later and bitch about the cost.

So, apparently, you can design the shear walls as independent 4' segments with OSB and anchors at every panel point (netting 2 per 4') and drive up cost. Or, you can have shear walls acting so the loads are transferred from one to the other per AWC (see Cadair)with the panels 'rotating' about their centroids and keeping the studs aligned. I've never been able to get a document that shows how this works. AWC says they're working on it.

Or you can have the latter without what I call boat anchors at each end like the current project I'm working on. It was designed by the same engineer who reviewed the first project. Their solution was to use 18"x18" thickened slabs, increase the slab from 4" to 6" and tie the anchors to the intersections and grab some of that 6" of concrete. I ran some numbers and what I call extra concrete means I could have had twice as many boat anchors at 6'x6'x2' for 25'x30' units.

It seems every engineer looks at each situation slightly differently. I can see that since forces can be resolved many ways. But the key, I think, is to optimize; optimize truss layout, optimize materials for shear walls, optimize whether boat anchors are less concrete than a 6" slab for a 12000 sf footprint, optimize the cost to pay for the design.

As an engineer once said, anyone can build a bridge but it takes and engineer to design one ... just barely.

Anyway, thanks again for your input.

CQ

 

[oldrunner]

jdgengineer: You are right about typical shear wall construction, but CurlyQue answered the question which hadn't been asked. Basically each four foot panel is a shear wall section in itself - only connected together with top and bottom plates which distribute the lateral loads to each wall individually. And that's why you would have to have holdowns at each end of the shear wall. This is similar to tilt-up concrete panels when they are not connected at the vertical joints.

It's kind of an expensive way to this because the sheathing could have been offset at each side so that the vertical edge nailing could have been connected to the adjacent panel.

As for the issue of buckling the studs, this is a simple calculation using first principles to check out the overturning forces. I would also be looking at sill crushing as well as stud capacity.

 

[RFreund]

Why would you want to use this method of separate panels as opposed to on long shear wall? You build the wall on the ground like normal construction and stand it up. Eliminates hold downs and is faster I would think. What is the reasoning for using this panel construction?

As for your analogies of the shearwall being similar to a cantilevered column or beam I'm not so sure about. Here is how I see it:
If you have a 10' tall story height with a 100' long wall your flanges=chords at the end of the wall and the depth of the beam/column is 100 ft however it is braces top and bottom so it is as if you have a plate on the top and bottom of your beam column to help prevent the web from buckling. So really it seems to be this 'weird' plate subject to shear and pinned on all side analogy. Or so I think.

EIT

http://www.HowToEngineer.com