Slab stiffness for lateral load analysis? 7

[PSR_1]

There is this argument around where I work which goes ' since post tensioned slabs don't crack during service loads, slab stiffness for lateral load computation can go as high as 0.5(eurocode, although it doesn't say for slabs) times the gross stiffness'. I amn't convinced with this statement as the contribution of the slab stiffness for lateral load resistance might demand special detailing. For RC slabs a value close to zero is usually used and am okay with that. But, I want to know your thoughts and practice regarding post tensioned slabs.

 

[hardbutmild]

HTURKAK,
I think that OP isn't really doing the analysis of the lateral system. Although I agree with almost everything said, may I ask what is the reasoning justifying the value q=4?
Sorry if this is a bit off topic.

 

[HTURKAK]

HTURKAK,
I think that OP isn't really doing the analysis of the lateral system. Although I agree with almost everything said, may I ask what is the reasoning justifying the value q=4?....

Dear hardbutmild, the SFRS is only Wall ( uncoupled Wall ) system and R=4 or q=4 complying with ASCE 7 and EC 8.

The ductile design of the shear wall for R=4 means that , eventually a hinge at the bottom of the Wall will develop at the reduced force level,.but will continue to deform while providing constant resistance up to some point and then eventually degrade to lower resistance level.

For the above mentioned case, PT slabs and columns shall offer deformation compatibility.

Another issue is the hinging at column connections ;the strong column-weak beam concept for MRF which shall not be valid for this case.The punching shear overstress failure which is the common reason for progressive collapse shall be prevented. I think the order of strengths shall be = punching shear strength (strongest), column strength and finally slab strength .The strong punching connection at slab-column and weak column will prevent sudden progressive collapse.

I would like to learn your thoughts for this concept.

 

[hardbutmild]

HTURKAK,
I agree, but I guess we design to different levels of ductility. As seen from the picture taken from EC8, at DCM q factor is 3,0 for wall systems.

I always avoid DCH because it requires very strict detailing provisions and where I'm from, I noticed that builders never follow the instructions like bending the stirrups 135 degrees (they bend them 90 degrees even when you draw 135 and write that it's important); they also put low ductility steel in critical areas and more bad things... That and the fact that a lot of the real structures are irregular in height is why I usually avoid high ductility structures. Basically, I doubt the workmanship and the ability to capture inelastic behaviour with elastic analysis.

EDIT: Note that EC8 doesn't require any building to be DCH, it leaves the choice to the designer, you can choose the combination of capacity and ductility. Both levels of ductility should be approximately of the same cost (this fact is taken from designers guide to Eurocode 8), but as they say it, DCM buildings show a little bit better behaviour for medium earthquakes, while DCH provides better behaviour at earthquakes higher than the ones in the code.
And that reminds me... whenever I required the highest ductility class of steel (steel class C in europe... B500C; the only one that's even allowed for DCH) the builders instantly called me to say that no such thing exists and that it's impossible to even produce steel of these characteristics (obviously a lie by them... but sitll, they certainly will not install it).
So, just be aware that these things might happen and if no cost benefit is achieved, go for a solution that's simpler to detail and construct. It's just my experience.