WUF-W Connection 4

[Rantamental]

Hi Guys,
Recently our company structural engineers team have designed a high rise steel structure with special moment frames system+ special bracing frames which boosted by rotational friction dampers and used WUF-W connection for connecting beam to column. According to ACI-358-16 (PreQualified steel connections) ,section 2.3.2a Built-up Beams Members, The web and flanges shall be connected using CJP groove welds with a pair of reinforcing fillet welds (min 8mm) within a zone extending from beam end to a distance not less than one beam depth beyond the plastic hinge.
Also according to WUF-W criteria, beams should conform the requirements of Section 2.3., And according to Section 8, the plastic hinge location shall be taken to be at face of the column, So protected Zone is a distance equal depth of the beam at face of the column and in this zone, Welding flange and web of the beam should be CJP+ 8mm (Min) fillet weld in both side of the web.
Our Design was similar to above descriptions completely, Also WPS and Welding Maps of shop drawings prepared true But unfortunately steel manufacturer contractor did not observe these limitations and has connected web and flange just by a 8mm to 12mm fillet weld according to web thickness in both side, therefore there is not any sign of CJP welding.
Now one-fifth of structure has been constructed and erected and we are looking for any criteria for acceptance of as built connections or any other implement works like repairing or retrofitting the connections to satisfy prequalified connections code.
Your prompt answer will be appreciated.

 

[sandman21]

I didn't suggest beefing up the weld. Perhaps you were addressing Deker there. I did find the parallels with the 75% provision in the bolted end plate section to be one of the more salient arguments presented here however.

You did suggest talking to the AHJ and trying to get a non-tested reinforcement accepted. One would assume you were talking about reinforcing of the welds. Regardless it has been suggested before to test larger fillet welds as a means to resolve the issue.

Facinating. Thanks for sharing that. Sans pre-qualification testing, if basically is what I sketched out conceptually.

Your concept relieves no stress from the flanges and the slotted web still has the same CJP requirements other assemblies have. It would be a stretch to apply the basic concepts.

There already exists a pre-qualified bolted flange moment connection. Seems to me that the bolt holes should be resolvable.

Are we in your hypothetical world or the current one? Does your hypothetical world have any pre-qualified? You cant go jumping from world to world without laying down some ground rules.

 

[KootK]

Are we in your hypothetical world or the current one? Does your hypothetical world have any pre-qualified? You cant go jumping from world to world without laying down some ground rules.

Fair enough. The ground rules of KootK Hypo-World are these:

1) No regulations impede the free use of engineering judgment in any way.

2) All information from all worlds, past or present, is available to inform judgment.

In this way, one could know of 358 and the research that underpins it without being bound to follow its requirements.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[Rantamental]

Dear Engineers
Thanks to all for your nice comments and beneficial negotiations
I have to say that we are consultant of this project and it is not possible for us to give or not a permission according to engineering judgment or theoretical sketch or traditional analysis! Any document should has a strong international reference and could be defend-able in any Court that may be formed in near future. I know, maybe (In real world maybe) that current fillet welds pass criteria of codes, but could you sign the permission documents and acceptance of current fillet welds with confidence with out any extra test? Also accept complete responsibility for this permission and defend it legally with out any testing ? Something about 1700 Tons of structure has been manufactured and erected with this deficit.

1) The particular girder detail shown has 30mm thick flanges, which are slightly thicker than the 1" limitation in AISC-358. Also, the clear span-to-depth ratio (2400mm/350mm = 6.86) appears to be slightly lower than the minimum ratio of 7 allowed for SMF. In practice, are these slightly out-of-spec values acceptable for prequalification?

Yes, you are true but our documents for designing has a supplement that allow us to use built up beam with flange thickness up to 30mm instead of 25mm.
Also span to depth ratio is 6.66 in critical girder and ratio of allowable value to current one is (7/6.66=) 1.05 and we were allowed to going beyond up to this. (5 Percent in all sections)
Be careful that there are a lot of issues which mentioned in AISC358 sections and commentaries that are not clear in theoretical( Limitations: Unit weight of girders, height of beams, span to depth ratio, beam flange dimensions, value of Cpr and Etc) and adopted from a lot of tests. Now, Could we get a unique result from all discussions or not?

 

[dhengr]

Rantamental:
We can NOT...
You get to pick what you can live with, prove to your satisfaction, and defend, we can’t/won’t do that for you. You certainly have been given a bunch of good, reasonably well thought out ideas and opinions, now it’s your job to apply your engineering judgement and experience to resolve the problem. You will have to be something of a diplomat to bring all the parties together, particularly the AHJ, and you hope they will be willing cooperative and constructive participants. When the fabricator, erector and GC consider the other possible alternatives, they should be able to find some money for testing, or whatever else is needed, to come to a reasonably acceptable resolution of this nasty situation.

 

[Rantamental]

Sure
I am with you and agree you
For us the most important issue is structure, safety and performance, so we offered just two following options
1- Test according to AISC 358 chapter k
2- change the members that do not comply criteria or change welds
It is clear that we can not accept any thing with out test, it is rule of engineering, try and error to find the way
Both of them are costly undoubtedly, but we could not find any better way, also I think the second one is impossible!!
Thanks to all for your helps and comments

 

[KootK]

One alternative that hasn't yet been broached here is performance based design. For a tall structure in which inelastic behaviour can be expected to be well distributed both laterally and vertically, I would expect joint rotation demand to be considerably less than the spec 0.04 radians. Even in high seismic regions, very tall buildings tend to be heavily influenced by wind performance concerns. The trick is that I don't know what, if any, lower value of joint rotation might obviate the need for testing. Perhaps other thread participants have insight to offer in that regard, even if it's to simply confirm that PBD is another dead end.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.