AS3600 - Cl 14.6.6 3

[GTD_18]

Hi All,

I was wondering what the "eng-tips" consensus is to cl 14.6.6 of AS 3600-2018 to allow for the effects of over-strength in the walls?
It may be a silly question but wouldn't this result in an unrealistic load allowance in most conditions - particularly if the walls are sized to limit deflection?
Is there other method that could be adopted to "make allowance" for the over-stength condition?

Any thoughts or comments would be appreciated?

Cheers

 

[IDS]

I don't do building work, but the principle seems sound to me.

If you want a section to fail in bending before it fails in shear, then you need to scale up the design shear strength to allow for the bending capacity being higher than the design value.

Doug Jenkins
Interactive Design Services

http://newtonexcelbach.wordpress.com/

 

[GTD_18]

Thanks for the comments IDS (apologies in the delay in response);

I agree the about the reasoning but i was wondering what the background of this clause is (i.e. why have a 1.6 factor rather than 1.05). Also i was wondering whether this assessment should be made against PhiVu_max over PhiVu?

Sorry if this is a daft question - i have been working on a legacy project utilising AS3600:2009 so i am just getting my head around the 20118 updates.

Cheers!

 

[Agent666]

That equation in AS3600 is fairly simplistic.

In NZ we would look at the actual overstrength capacity, this is based on an assessment of the wall moment strength based on replacing f'c with f'c+15MPa to account for the average concrete strength longer term, and replacing f_y with the 95% characteristic strength of the reinforcement (for 500MPa reinforcement this is 1.35x500 = 675MPa). This overstrength moment capacity is intended to be the upper bound moment strength/capacity. Due to the formation of ductile mechanisms, the overstrength axial load also might need to be considered in making this assessment, further increasing the overstrength capacity. For example in coupled walls with diagonal reinforcement in coupling beams, in yielding the diagonal reinforcement in coupling beams, the overstrength demands on the coupling beams needs to be assessed, and this increases the axial design actions in the adjacent wall piers.

This is where the simplistic 1.6 factor comes from, it is the ratio of M[sub]O[/sub]/ ØM[sub]N[/sub], where M[sub]O[/sub] is the overstrength capacity.

The point is depending on your particular wall arrangement and the axial load and where you are sitting on the M/N interaction curve for your wall the 1.6 factor could actually be anywhere between 1.3-2.0 times the ultimate moment capacity (ΦM[sub]N[/sub]). So hard coding in 1.6 could in some scenarios actually under estimate the overstrength shear demand. In NZ we would take a phi of 1.0 for the shear check as the capacity design approach (i.e. upper bound moment demand/capacity) ensures the load cannot be higher. So if in the AS3600 standard if you are not doing the same thing, and phi = normal value, it might cover the worst scenarios.

Hopefully that clarifies the 'why is factor 1.6 and not 1.05' question?

 

[captain_slow]

Agree with Agent666 regarding the origins of this clause and its intent.

My first thought was that beyond the purpose of this clause it is also just a daunting calculation to perform for every wall in a reasonably large project - you essentially have to compute moment capacities for multi-reinforced sections with many many layers.

Now that live projects are ticking over into the new code (I believe AS3600-2018 wasn't referenced by BCA until last year) I actually sat down and typed up an Excel macro that performs this calculation. My two cents on this clause is that for great majority of the walls that I have checked this far my V* values from analysis were almost always higher than this clause requirement.

My take is that this is a calculation-heavy clause and it doesn't have have as much impact on design as the rest of section 14.

 

[GTD_18]

Thanks for the feedback – great responses. After a little bit more research I found that the 1.6 factor is based from a recommendation within the Priestley 2007 “displacement-based seismic design of structures” – pg 170.

So, going back to clause 14.6.6 (and hopefully not pushing may look); I assume that when we establish Mu we would adopt the axial load of G+0.3Q and the M* and V* would be a function of the EQ response (i.e. G+0.3Q+EQ).
So, how do we accommodate walls or frames with dead load (G) shears built in? for example, if we have a frame system that has say 500kN shear due to the dead load do we also scale up this shear too or do you isolate the shear due to the “earthquake design action” as noted in the code?

Very much looking forward to the commentary to this code update - even a preview to section 14 would be helpful.

 

[li0ngalahad]

Normally walls with very small applied moment, have a much larger moment capacity, especially when you need to comply to the minimum reinforcement specified on 14.6.7, and this will sometimes lead to almost undesignable overstrengths. For example let's say my wall has a Mu=5000 kNm but a M* of 500 kN (Mu=10M*) and say V*=1000 kN, I would need to design my wall for 16V* or 16000 kN!!!
My approach, if I am using mu=2 and Sp=0.77, would be to assume the wall's moment capacity to be fully utilised with M* equals phiMu, so M*=0.8Mu, therefore I can design my wall shear for 1.6/0.8 V* which means 2V* (double the actual shear), which I would think it's a fair shear overstrength and would match the shear I would get if I used mu=1 and Sp=0.77 (which would not require any overstrength).
Anyone else adopting a similar approach?

 

[mrlm]

Hi liongalahad,
The draft amendment 2 altered the equation:
phi.Vu > min [(1.6 x Mu)/M* ; (mu/Sp).V*]

This means the upper limit (if using mu=2 and Sp=0.77) is now 2.6 x V*

 

[li0ngalahad]

Great, this way they put an upper limit to this requirement