Extended Shear Tab Connection

[ship4885]

I'm reviewing shop drawings for a building our company provided the engineering services for. The detailer has provided extended shear tabs for all direct shear connections at beam/girder webs as well as column webs. Where we have a cantilever, there is a weak axis moment connection. It is a direct flange weld to plates welded to the flange and web of the column. The shear tab is welded to both the column web as well as the plates, so they are scting like stiffners to the connection.

My question is where to take my eccentricity for the bolts on this connection? If these stiffners restrain my shear tab from rotating, can I assume my eccentricity is from the centerline of the cbolts to the toe of the flange, or must I still consider eccentircity to the web of the column? The design checks outlined in the 13th edition do not note specifically if your eccentricity changes due to a stiffened connection.

 

[vandede427]

I just attended a seminar about this. Essentially the bolt group on the attaching beam web to the extended shear plate is so rigid that the tab is almost an extension of the attaching beam's web. Therefore any eccentricity on the attached beam is no different than if there was a double angle clip connection.

 

[ship4885]

I think you possibly missed what I am saying. I agree, if my bolt group is designed for the eccentricty from the cneterline of the bolt group to the column web, and my plate strength is limited as outlined, then yes, you are basically extending the web.

But I am asking, if this shear tab has stiffners on top and bottom, must I design for the eccentricty to the column web, or the toe of the flange.

 

[ship4885]

Here is a sketch for any clarification.

 

[WillisV]

Column web, unless you specifically design the column for the moment induced due to the eccentricity from the end of the siffeners back to the column web - in that case you can design the bolts only for the eccentricity from the end of the stiffeners to the bolt line.

 

[hokie66]

I think it is just a web shear splice. The moment is taken into the column by the flange plates. Designing the web shear bolts for the same moment is not necessary.

 

[mudflaps]

Why not check it both ways and see how much difference there is. There might not be enough to be worrying about. You'll sleep better if your design goes to the column web and the cost difference will most likely be lost during bidding.

Old CA SE

 

[WillisV]

hokie66 - I believe the configuration being discussed is simply an extended shear tab with horizontal stiffeners at the top and bottom within the width of the column - they don't actually extend out to engage the beam.

With this type of extended tab connection, the bolts could be say 6" out from the face of the column where the horizontal stiffeners end. If the stiffeners end 7" from the column web, you have two options. Either 1, design the column for no moment and take an e of 13" in the bolts (thus assuming the pin is at the column web). Or 2, design the column for an e of 7", and the bolts for an e of 6" (thus assuming the pin is at the end of the horizontal stiffeners).

 

[hokie66]

Willis,

His picture shows the beam flanges prepared for welding, so I think intends a welded moment splice at this point.

 

[WillisV]

Hokie - Ahh...looking at the sketch would help...didn't realize this was a full moment connection with CJP flanges. I agree with you then, just design the bolts for the direct shear as the column and stiffeners are already designed for the moment.