Steel Frame OMF to SMF 1

[YoungGunner]

BTW, I'm not looking for solutions or alternative ideas - simply asking if this logic checks out.

The height of the frame is around 45ft. It exceeds the 35ft limit for a multi-story steel frame for OMF. It would change the architectural look of the designer's ceiling to do a SMF all the way up and deal with bottom flange bracing on the top beam at the roof. We want to avoid Simpson's specialty product that doesn't require bottom flange bracing. So, can we do an OMF on a SMF (we can do bracing at the floor level easily) that share the same columns? Obviously the R of 3.5 tracks into the SMF, but the goal here is to fall underneath the provision of a 65ft single story OMF by having it stack on an SMF, and avoid bracing on the top beam. Does this logic check out?

 

[KootK]

I refer to the commentary of AISC 341 for my thought process for OMF,

I see. I mistakenly thought that you were suggesting that yielding might occur anywhere within any frame element. What you were really suggesting is that yielding is permitted to occur:

1) Generally adjacent to the same joints that it always does for IMF/SMF but;

2) Within any of the members coming into those joints, be they beam or column.

That I certainly agree with.

Coming up with sufficiently precise language for these seismic chats is a real challenge.

 

[JoshPlumSE]

In general we do not allow hinges to develop in our columns. That's the whole "strong column / weak beam" thing.

I'm probably picking nits here, but there is a common misconception in this statement. Hinges form where they want to form. It's not under the complete control of the engineer. The actual excitation of the earthquake may be different than anticipated, the non-structural elements may change the behavior and so forth.

In SMF frames, we do expect a decent amount of inelasticity in the columns. That's just what the test data shows. I believe that our Strong Column / Weak Beam provisions (which I don't think are named that anymore) are an effort to reduce this to a point where the columns can still support their gravity loads post-event. But, you will likely still get some amount of hinging in the columns. I believe the name change for those provisions were made because it was felt that it could be misleading.

 

[jerseyshore]

driftLimiter, I was just responding to the comment above regarding contacting the city to see if they would approve it. I agree that it doesn't matter, that we have to design to code, I'm just saying asking the town isn't something I would even consider for something like this. Building officials around here rarely have structural comments. Maybe 1 in every 20 jobs and then it's usually just asking for the wind speed or something trivial.

YoungGunner, what did you end up doing?

 

[KootK]

Building officials around here rarely have structural comments.

That's also true of the markets in which I've worked for the most part (Wisconsin / Alberta).

I do some high end residential in California where they have folks who are sort of "professional reviewers". I'm able to pull the wool over their eyes on pretty much anything unless it's an obvious violation of a code clause interpreted to the letter. They're quick to give me flack if I forget to label a shear wall. They won't even notice if the overall shear wall arrangement does not satisfy equilibrium. If you've got Jack Mohle reviewing your PBD design of a skyscraper, I assume it's a bit better. That said, I've seen some deeply flawed coupling beam detailing slip by those guys too.

 

[KootK]

I'm probably picking nits here, but there is a common misconception in this statement.

Looks like I may be in for some re-education after all. I feel that my original statement had no misconceptions embedded within it and, rather, it is your latest post that is in error. I'll just go ahead and cry foul as I see fit and we'll take it from there.

Hinges form where they want to form. It's not under the complete control of the engineer.

1) We DO NOT control which hinges will develop first throughout a building with respect to plan location or vertical location in tall buildings.

2) We DO control the hierarchy of failure of the various failure modes at any given joint. This is the very crux of the capacity design method that governs our routine seismic design work.

The actual excitation of the earthquake may be different than anticipated, the non-structural elements may change the behavior and so forth.

The original goal of capacity design (Park & Paulay et all) was to deliberately develop a design method that would be insensitive to such sources of variability and uncertainty.

That such sources of variability and uncertainty exist does, in no way, invalidate the strategy of purposely designing members and joints to yield and fail according to a desired hierarchy.

I believe that our Strong Column / Weak Beam provisions (which I don't think are named that anymore)

What are they named now? I don't want to be running around here sounding out of touch.

Strong Column / Weak Beam provisions...are an effort to reduce this to a point where the columns can still support their gravity loads post-event.

When I consider Strong Column / Weak beam, I visualize the sketch show below, taken from one of the NEHRP seismic design guides. The goal is to prevent all of the columns within one story from forming plastic hinges that would cause an individual story to pancake. This is critical during a seismic event, not just after a seismic event.

There... now you go.

 

[JoshPlumSE]

KootK -

We DO control the hierarchy of failure of the various failure modes at any given joint. This is the very crux of the capacity design method that governs our routine seismic design work.

This was my whole objection. We do not CONTRL the failure method at any given joint.... We merely bias our design using these ratios to ensure better overall system ductility.

I understand I was picking nits. The general concept you described was correct.... But, these provisions do NOT prevent hinging from occurring in columns. Rather it is intended to make it so that beam hinging dominates in the frame (especially at multiple levels) and leads to great system ductility. Even though you may still get significant inelastic behavior in your columns, ideally through panel zone yielding. But, also through flexural yielding in the columns.... just hopefully not at multiple levels in a way that would cause system failure.

The image in your last point is more about soft story issues than it is about moment ratio. I should point out that if those columns only hinged at the base then this would actually be FINE for a SMF. In fact, I think column base hinging is a reasonably common form of column inelasticity.

My Source:

The strong-column weak-beam (SC/WB) concept is perhaps one of the least understood seism provisions in steel design. It is often mistakenly assumed that it is formulated to prevent any column flange yielding in a frame and that if such yielding occurs, the column will fail. Tests have shown that yielding of columns in moment frame sub-assemblies does not necessarily reduce the lateral strength at the expected seismic displacement levels.

The SC/WB concept represents more of a GLOBAL FRAME (my emphasis) concern than a concern at the interconnections of the individual beams and columns. Schneider, Roeder, and Carpenter showed that the real benefit of meeting the SC/WB requirements is that the columns are generally strong enough to force flexural yielding in beams in multiple levels of the frame, thereby achieving a higher level of energy dissipation.....

It should be noted that compliance with the SC/WB concep and equation 9-3 gives no assurance than individual columns will not yield, even when all connection locations in the frame comply..... Nonetheless, yielding of the beams rather than the columns will predominate and the desired inelastic performance will be achieved in frames sized to meet the requirement.

Note: After reading this commentary, I realize that the reason why this stuck in my brain so much is because I was having a conversation some years ago with Charles Roeder at an AISC committee meeting where I used the term Strong / Column Weak beam as a way of preventing hinging in the columns. Very similar to what you said. And, he replied with a comment that was VERY similar to what is given above from the AISC 341 commentary.

 

[sandman21]

Just read the commentary and research on panel zone shear behavior and yielding in the zone. Refinement to this will continue as its a location with good ductility and performance but which we do not full know the limits. Most of the code are prescriptive limits based on past performance of a given system. Low-rise OMF have shown sufficient performance when limited in height, weight and plan dimension. Old OMF buildings in San Francisco performed better in Loma Prieta than newer frames in Northridge. Nearly every connection was OMF reducing the demands and increasing system performance. I am sure we could design a building as OMF to the heights of SMF but the number of connections and limits on the building would be to costly.

 

[KootK]

@JoshPlumSE: yeah, that's pretty damn nit picky. So, basically, it comes down to nothing more than this:

My original response + acknowledgement that it's a whole story thing + acknowledgement that our "protection" of columns is imperfect.

That's not a misconception. It's just me not bothering to type drill down into every possible nuance of the thing.

In general we do not allow discourage hinges to develop from developing in our columns. That's the whole "strong column / weak beam" thing.

Good?

 

[X4vier]

YoungGunner

Can you share your final solution?

 

[X4vier]

I've seen that multistory SMF buildings don't use Moment connections at the roof level only at floors.

Can I just extend the end of the columns and bear a simple span beam on them?

 

[X4vier]

Please I need your opinions!!!

 

[Luceid]

X4vier, your drawing above seems to show a gravity frame on top of a moment frame. If you just wanted the SMF at the bottom to act as a lateral element, you can do whatever you want above it. However, this would be assuming that the columns above your SMF dimension would not be serving the purpose of transferring lateral load from the roof down to that moment frame. As long as you don't expect that upper pinned frame to be a lateral element in any way, it's just a gravity frame and has nothing to do with OMF/SMF, so you can't count on it as a seismic element (which will pretty much happen if you pin those beams anyways, I think). But there's a case where you could theoretically do this if that's what you intend for it to do.