Horizontal Trussing Threshold for Roof Diaphragm 3

[mes7a]

Hi,

How rigid must be the horizontal trussing in a roof before it can be considered a full fledge diaphragm? When the structure was constructed. They use 0.5x0.5 meter column from foundation all the way to third floor (it was designed for 4 storey).. but we decided now to build up to 3 storey only and plan to use thin metal roof to shed rain at gutter at side.

I know there must be a threshold in the horizontal trussing before the huge columns can become a diaphragm. For example. If you merely use a pole connecting the columns.. it can't be considered a diaphragm. Is it?

Also must the rafters be straight horizontal.. how about a bent gable rafter with apex at middle.. can a bent rafter create a diaphragm.. again what is the threshold of the rigidity? The pole is also rigid.. must it be certain strength compared to the huge columns?

Thank you.

 

[KootK]

The secondary beam reaction will be cusing a crack on main bar if the main bar can not resist it and if it has not been provided with extra stirrups on either side at a distance "d" and in addition as one of the sefians posted crank or bent up bar is provided.

I believe that the referenced cracks are cracks in the concrete surrounding the rebar rather than cracks in the rebar itself. Rebar yielding should precede rebar rupture and concrete rupture should precede rebar yielding.

hopefully the concrete can carry some load.. would it.. how do you analyze this in your design?

It would. I analyzed it using the hanger steel provisions or, in unusual situations, with strut and tie.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[mes7a]

I believe that the referenced cracks are cracks in the concrete surrounding the rebar rather than cracks in the rebar itself. Rebar yielding should precede rebar rupture and concrete rupture should precede rebar yielding.

Yes. This is what I need to know. Last night I couldn't sleep because pondering if rebar can also suffer brittle shear failure or would it yield first then break. But then if you hammer a rebar on the side. It surely can break before yield.. no?

So the Indians may be saying that without the stirrups close to the interface.. the bars may bend and then yield and then break. Judging by his tone I thought he meant it would just break brittle.

But then for higher yield rebar. It is more brittle.. it doesn't go into hardening and strain is less. In this case, the bar can literally break.

By the way. Before we wrap up the thread.. there is something the indian said but I didn't follow up 2 years ago and don't understand. He said:

"With beams framing plan as per sketch, irrespective assumptions made for design, this system is really a grid beams with ends of some beams rest on columns[ unyielding support] and some on beams [ yielding supports, since at the point deflections will occur]. This will lead to additional shear in 6m span beams resting on beams. "

What is other terms for "unyielding support" and "yielding supports"? Is he talking about rebar yielding or support conditions? He refers to yielding supports connected to deflections.. why would it be have additional shear in the secondary beams resting on primary beams? What is the equivalent concept in ACI? They may use other terms in their indian codes.

Many many thanks.

 

[KootK]

But then if you hammer a rebar on the side. It surely can break before yield.. no?

It shouldn't unless it's extremely cold, overly cold worked, embrittled via improper welding, containing manufacturing defects etc...

"unyielding support" and "yielding supports"? Is he talking about rebar yielding or support conditions?

Support conditions. "Stiff supports" and "felxibile supports" would have been more apt terms.

He refers to yielding supports connected to deflections.. why would it be have additional shear in the secondary beams resting on primary beams?

I'm afraid that I don't understand the referenced statement. Monolithic concrete structures are very indeterminate. As such, it's always tough to know how much load is going to the various elements.


I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[mes7a]

I'm afraid that I don't understand the referenced statement. Monolithic concrete structures are very indeterminate. As such, it's always tough to know how much load is going to the various elements.

His complete sentences is "This will lead to additional shear in 6m span beams resting on beams. Since length of both beams together is 12m, bottom bars may have been lapped at the junction. This lap shall be not less than Ld/3 or depth of beam, whichever is larger. This may be one of the reasons.
Cracks may occur anywhere where the deficiency is."

He doesn't know I use 12 meters long bars so there is no lap. I think he meant there may be more shear if the lap is not Ld/3 at critical portions owing to insufficient development length.

Anyway. I think that's it. I'll pause, ponder and reflect every word you said in this long thread and others. You patience is top notch. I don't know how to thank you. But thanks a million anyway for the many responses in this thread. You are incredible Kootk. Someday if you can reveal to us your real name, better so we can nominate you as educational system official! Lol.

 

[mes7a]

You can check the capacity using strut and tie procedures applied to the opening and closing models that I sketched above. Another way to look at it is to flip the model upside down in which case it becomes much like a roof level slab to column connection. The 270 kNm that I mentioned earlier was a ballpark number based on having to yield 8-20M column bars to over strength at a lever arm of 300 mm. I'd come up with a detailed estimate of your own before investing too much time on it.

Btw.. Kootk.. something important to ask...

What do you mean by "lever arm of 300mm"? Did you measure it from rebar to rebar at sides? Again, the columns at the sides are 0.5x0.4 mtr.. not 0.5x0.5 mtr (which is only used at center).

Please just focus on the C1 which is used on all the sides and corner except one.

Where did you come up with 300mm?

I have spent the year 2013 in manual computations of columns and beams to verify loadings. In the following:

You can see only half of the bars are used in moments in tension side (the compression side bars are to add to the compression block).

So when the column footing connection developed plastic hinges.. it is only one side of the column that would yield.. so since there are only 6 bars that would yield.. then one has to calculate the contribution it alone right? (because you mentioned above that "having to yield 8-20M column bars to over strength at a lever arm of 300 mm".. why do you have to use all 8-20M? or all 6-20M since we are talking of the 0.5x0.4 mtr C1)?

Also would it even get to yield before the compression block on the opposite side of the tension crushes?

My designers still told me the punching shear alone is enough without having to think about flexural capacity of the foundation.. so I'll plan to manually compute for the polar moment of inertia and see if the flexural capacity would be enough. Don't worry. I won't use it on new structure but just to verify my own existing building. Thanks so much again.

 

[mes7a]

Haynewp

I don't know what codes don't forbid OMF in high seismic.

I don't think using R=1 necessarily means an elastic design. There is a method to keep critical structures close to elastic in an EQ, but it involves a time-history suite and R=0.85 if I remember correctly. I will try to dig up the reference.

I read R=1 means there is no ductility or sorta.. I tried to find the reference about this method you mentioned to keep critical structures close to elastic in an EQ that involves a time-history suite and R=0.85 but couldn't find it. If you dug it up.. please share it (ok in other thread as this is quite long already and I know need to be closed).

I just want to know how big are the moment demands in columns for them to perform elastically and how big they should be.. and what kind of analysis it is. Thank you.

 

[mes7a]

You can check the capacity using strut and tie procedures applied to the opening and closing models that I sketched above. Another way to look at it is to flip the model upside down in which case it becomes much like a roof level slab to column connection. The 270 kNm that I mentioned earlier was a ballpark number based on having to yield 8-20M column bars to over strength at a lever arm of 300 mm. I'd come up with a detailed estimate of your own before investing too much time on it.

Oh Kootk. I got confused for a while as I let every word you said sink in. I thought your ballpark number estimate was based on 20-20M (which I erroneously read)... but it's 8-20M.. so you are still right (I was hoping you were wrong somewhere and the footing foundation flexural transfer can be done. Although your lever arm of "300" may really be 400. I'd review the computations in our favorite textbook next week.

Whatever. If the axial load is below the interaction diagram balanced point. The column base would yield even when moments is not so much.. this means the footing-foundation connection won't be engaged as the column would fail already first (yielding to breakage). And when it is above balanced point.. the concrete would crush first... therefore consider the column flexural as possible limiting factor that can already fail before it can engage the flexural overstrength yielding lines of the footing column interface. Here there are 2 parts that are yielding independently (column tension side and the connection flexural side). In the future if found a good reference about this. Please share it.. thanks a whole lot.

 

[KootK]

Yeah, my ballpark estimate of the column flexural over strength was just that: a ballpark estimate. I used my judgement to guess at the effective lever arm and the number of tension bars participating. For real checks, you'd want a better, independently calculated estimate.

For a special moment frame, fixed based columns have to be able to yield at their bases without brittle shear/crushing failure. As such, the footing would never be prevented from experiencing the column over strength moments.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[mes7a]

Yeah, my ballpark estimate of the column flexural over strength was just that: a ballpark estimate. I used my judgement to guess at the effective lever arm and the number of tension bars participating. For real checks, you'd want a better, independently calculated estimate.

For a special moment frame, fixed based columns have to be able to yield at their bases without brittle shear/crushing failure. As such, the footing would never be prevented from experiencing the column over strength moments.

If the axial loads of my column bases at foundation is very low or a bit far from the balanced point. Low moment or low seismic activity would already yield it. This means the footing would take lower flexural strength to experience the column overstrength moments. Right.

But then when my column bases yielded at low magnitude 6 earthquakes and the foundation remain intact.. because it experiences lower moments.. remember the reason for it experiencing much less moment is because when the column tension side yielded, the compression block is not at maximum compression (less moments) and remembering the compression side block of a column section also participates in moments. Therefore I think a good strategy is to determining where is the axial load of my existing structure to see how far or near or if it is in the balanced point isn't it.. so moment capacity would increase.. therefore if my present 2 storey with roofdeck is much lower than the balanced point.. adding light roof won't change it much.. then have to finish the original 4 storey concrete building.. perhaps the axial loads would be in the balanced point or near it.. is this a good way to think? At balanced point moments with higher moment capacity. if the Big One (Earthquake once in 200 years) really came.. and the footing separates from lack of complete moment flexural overstrength.. well it's as bad as the building experiences complete mechanism.. noting the soil/rock underneath is much harder than designed (existing foundation is 3 times bigger.. the designers admit.. so in the even the footing separates.. the rock below has enough bearing pressure to support the separated column base... the point is.. in both situations.. they would both be useless (separated column bases or complete hinging mechanism formed)anyway after a major seismic activity.. so need to demolish and building new one. What you make of all this especially making sure the building axial loads need to be checked and be made close to the balanced point of the interaction diagram for more moment capacity and seismic resistance until footing separate or complete hinging mechanism?

 

[mes7a]

to continue.. unless you are saying that even if column moment is much lesser than the balanced point, if the seismic demands push the moments to the balanced point.. the footing would still experience that moment (of the column base as a whole) even if the tension part of the column is already yielding? In other words, the column base foundation flexural overstrength capacity is depending on the column moment as a whole and not on the tension part yielding moment of the column.. isn't it. Point is if the tension part yields, the column can still bend up to the balanced point? This is your point isn't it.

So irregardless if the axial load is really in the balanced point or much below.. the column moment would still reach the balanced point and this is where your column base footing flexural overstrength comes in? I was thinking in last message that if the moment is much lower than balanced point.. the column moment can't be push higher to reach balanced point moment before the yielding tension parts break apart, you think this is possible too?

 

[KootK]

is this a good way to think?

Mostly. I see two possible flaws in your logic:

1) you seem to assume that your existing columns were not properly designed to form plastic hinges when the structure was a four story building as originally planned. How do you know that is the case? If your columns are fine, you may well be penalizing your building -- and your sleep -- unnecessarily.

2) we design seismic plastic hinges to reach a certain capacity and then maintain that capacity while not resisting additional load. This is very different from a member or joint losing all load resisting capacity after the formation of plastic hinges. This needs to be kept in mind when considering the kinds of footing and column "failure" that may be acceptable.

You've asked for additional references several times now. And, in my estimation, the aspect of seismic design that you struggle with most is the general philosophy of capacity design. To that end, I recommend acquiring and digesting this reference: Link. It's one of the source documents that underpins much of our modern seismic design methodology. Normally, I would hesitate to suggest that someone read an entire, complex, out of date textbook in order to answer their eng-tips question. In your case, however, I know that you won't be deterred by the effort implied. And I'm confident that the exercise would answer a lot of the questions that nag at you.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[mes7a]

I have read that book before but ok I'll read it again for a week or so.

My point is simply this. Imagine you have a footing and the column is just half meter high and you connection a hydraulic bender to the column. It has zero axial load. Let's take the following case.

At zero axial load. Moment capacity is 180 ft.kips.
If you bend it with the hydraulic til it reaches 180 ft.kips, the rebars yield...

If you continue the hydraulic bender, can you push the moment to reach the balance point at 300 ft.kip? or would the yielding bars break halfway?

If it breaks halfway or before reaching the balance point.. then the footing flexural overstrength is dependent on the above.. isn't it.. ok.. this is my last question before I read the whole book again. Remember my designers told me their job is just to output reinforcement ratio in etabs.. they don't understand even the interaction diagram manual computations in any details or even plastic hinges.. when I told them about plastic hinges.. they said concrete is not made of plastic and just laugh at me. so I know I must seek others but need to know this basic before going on. Thank you.

 

[KootK]

I have read that book before but ok I'll read it again for a week or so.

I'm simply trying to help. I'm not attempting to assign you redundant homework. Obviously, I have no way of knowing what you have and haven't read unless you compile a list for me.

If you continue the hydraulic bender, can you push the moment to reach the balance point at 300 ft.kip? or would the yielding bars break halfway?

In a properly detailed SMF column, neither of these things should happen. This seems to be one of the fundamental aspects of capacity design that may be eluding you.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[mes7a]

In a properly detailed SMF column, neither of these things should happen. This seems to be one of the fundamental aspects of capacity design that may be eluding you.

In a "properly detailed SMF column".. that's right.. but my column is not properly detailed SMF. The following pic shows why.

10 gallons of pure epoxy injected:

It is honeycombing in the column because of poor workmanship by incompetent contractors. The designers told me to inject 10 gallons of pure epoxy. In the thread

http://www.eng-tips.com/viewthread.cfm?qid=341916,

everybody here is convincing me the low modulus epoxy won't take much load and can't be used a repaired material. But the designers won't listen because they forgot the concept of modulus of elasticity or stress-strain so can't understand what BARetired is saying. In the following quote. BARetired is commenting I have loss of axial capacity of about 1000kN. BAretired said:

" Just a comment on your earlier calculations (shown in blue):

Now to compute for the load reduction carried by the epoxy filling. I'll use strain of 0.0005 or load in the elastic range.

from steel strain=0.0005, Modulus 29,000 ksi
stress = strain*modulus = 14500 psi or 99973.98 pascal 100MPa or 100*106 pascals.

from concrete strain 0.0005, Modulus 3604.996 ksi (let's call it 3600 ksi)
stress = strain*modulus = 1802.498 1800 psi or 12427.79 pascal 12.4 MPa

from epoxy strain 0.0005, Modulus 450 ksi
stress = strain*modulus = 225 psi or 1551.32 pascal 1.55 MPa

Column is 0.5x0.5m, the 0.2x0.5 section was replaced with epoxy, remaining 0.3x0.5 section with concrete. In other words, 33% 40% of section replaced by epoxy.

steel area of 12 20mm bars (for concrete section) = 0.003769 mm^2 3600mm2...in Canada, 20M bars have an area of 300mm2, could be different in Philippines
steel area of 8 20mm bars (for epoxy section) = 0.002513 2400 mm^2

For load carried by concrete section (0.3x0.5 of column) with 12 bars of 20m steel
P = Fc(Ag-As)+Fs(As) = 12427.29(0.146231) + 99973.98 (0.003769) = 2194.13 KN
12.4(300*500 - 3600) + 100(3600) = 2175 kN.

For load carried by the epoxy section (0.2x0.5 of column) with 8 bars of 20mm steel.
P = Fc(Ag-As)+Fs(As) = 1551.32 (0.097487) + 99973.98 (0.002513) = 402.4681 Kn
1.55(200*500 - 2400) + 100(2400) = 391 kN

For load carried by entirely concrete(0.5x0.5 of column) with 20 bars of 20mm steel
P = Fc(Ag-As)+Fs(As) = 12427.79(0.243718) + 99973.98(0.006282) = 3656.913 Kn
12.4(500*500 - 6000) + 100(6000) = 3625 kN

Loss of axial load due of the epoxy is
P(all concrete) - (P(concrete)+P(epoxy)) = 3656.913 - (2194.13+402.4681) = 1060.3149 KN
3625 - (2175 + 391) = 1059 kN

Note: The above calculation assumes uniform strain throughout the column. If a transformed section is used, the centroid of the combined section would shift toward the concrete portion. That would cause bending stress in addition to axial, so the condition is likely going to be worse than calculated."

Kootk. Have you seen construction this bad? In our country. Injecting epoxy is the normal even on column bases on 30 storey high rise structure that has 2 meters size honeycombing. BARetired said it's so dangerous.

It's good the honeycombing is in my tension side of the eccentric front column. If it is on the compression side.. it is so scary.. but reverse cyclic loading worries me. This is the main reason why I don't want to build the 4 storey.

Have you seen one building with 10 gallons of epoxy injected and massive carbon fiber installed and even pedestral retrofit done? Or have you seen worse.. lol..

This is the reason why Im learning what a properly detailed SMF column should be.. so asking if it is possible the yielding bar breaks before the moment reaches the balanced point.. etc. Thanks.

 

[mes7a]

Kootk... maybe can use your almost genius muldisciplinary knowledge one last time to bear on this problem that is only local to our country and yet completely ignored here in my country.

This concerns the holes in the column replaced with soft material (epoxy). This is our only solution in our country for most incompetent concrete pouring. The location is in red in the following building plan (base of center front column).

from side view the hole looks like this

The column involves is the 0.5x0.5 meter

I could have created a new thread of this but those who is not familiar with my structure may be confused as to its location.. for example BARetired who scrutinized it 2 years ago thought it was at midspan.

But the above shows it is actually in the level of the tie beam at front center column.

Both my contractor and designers are forcing me to accept it because they said it would perform like concrete.. but as the multidisciplinary team here emphasized. It functions like soft material.

My question is.. with the beams connected to it in above at at sides. Would the front column still bend via the hole with softer material epoxy (the bending stress BAretired describing below)?

This shows the front center column has perpendicular beams connected to it continuously from center (in the picture we haven't put the extra bars yet)

Let me quote what BAretired said about column with epoxy hole generally:

"If a column is hinged top and bottom and compressed with axial load P, the stress is uniform at every cross section, namely P/A.

If a rectangular notch is cut out of the left side of the column at mid-height, the centroid moves to the right at the notch. The axial load falls to the left of the centroid and the notched section will move right relative to the hinged ends.

If the notch had been filled with material with low E, the behavior will be similar, but the centroid will not shift as far so the filled notch will not move as far to the right as the unfilled notch because the fill is carrying some stress but not as much as the concrete.

You asked why the concrete would crush. You do not have a perfectly rectangular notch across the full depth of the column. What you have shown on your sketch looks like an irregular cavity. Some of the concrete extends to the left edge of the column and that is the part that would crush first.

If axial stress exceeds bending stress, there is no tension on any part of the cross section, simply variable compression with maximum value on the left and minimum on the right.

At a strain of 0.0005 and a depth of only 4" of epoxy, I would not expect enough strain in the column to cause the kind of problem described above. But if the column is carrying its full design load and the epoxy thickness is 12" instead of 4", I would have serious concerns about the safety of the structure."

Kootk.. strain of 0.0005 is in the elastic range. This is the reason I don't want to top off the full 4th storey because I'm not sure of the behavior and no one locally can scrutinize this level of details. So want to make it just be within the elastic range.

But during seismic reverse cyclic loading, the structure would move back and forth.. wouldn't this cause the bars in the holes with epoxy to yield because it's taking mainly the load from the softer epoxy? I've been thinking of this non-stop for the past 24 hours after reviewing that thread again. What is your experience of this.. what do you make of it? I did the following retrofit to fix the epoxy retrofit. But if the columns bends.. the concrete surrounding it may not restrain it (because the bars circling it may not be enough). Any other solution to maybe retrofit the 2 retrofits again?

retrofit rebars:

retrofit concrete:

Thanks a lot for all the multidisciplinary comments!

 

[haynewp]

Section 4 here has a procedure, it is using R=1 as part of the requirements. I must have been confusing R less than 1 with something I saw in AASHTO recently on connection forces under seismic.

https://www.wbdg.org/ccb/DOD/UFC/ufc_3_310_04.pdf

 

[KootK]

I struggle to think of a suitable repair strategy mes7a. I agree, the construction quality is terrible and the existing repair quite unsuitable.

One could perhaps encase the entire first floor columns in new concrete and rebar. That would, of course, change many of the seismic parameters that we've been discussing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[mes7a]

I struggle to think of a suitable repair strategy mes7a. I agree, the construction quality is terrible and the existing repair quite unsuitable.

One could perhaps encase the entire first floor columns in new concrete and rebar. That would, of course, change many of the seismic parameters that we've been discussing.

I have personally check and hammer each ground floor column. The others are ok. The reason the front center is problematic is because if you can see the layout below (arrow in red).. the tie beams are slant and the carpenters couldn't do proper formworks without leak and some rebars congestion there caused it.

I was hoping the column can act as gravity column only and the 8 other moment resisting columns are the main lateral resisting columns.. as you said to make the gravity column ride along with others.. Also someone mentioned in the thread 2 years ago that:

"Is this ONE column in a series or group of columns that contribute to the seismic resisting frame? If so, would it be possible that the remaining columns could support the lateral loading? This ONE column would only have to be retrofitted to support the vertical loads and detailed in a manner that it only 'go along for the ride' in a seismic event."

Btw.. our rebars here is 314 mm^2 for 20mm.. not 300mm^2 for 20M (which is in Canada). At bar strain of 0.0021.. the 8 rebars (at side only of the 20 rebars) have axial capacity of about 1000kN (i'm thinking how to convert it to moment capacity).. my building axial load above that column is only 700kN so the present section with concrete and epoxy and rebars have capacity of 2000kN. I'm still manually familiarizing myself with the interaction diagrams now reviewing what I learnt 2 years ago to see the composite behavior of epoxy and concrete. I'm doing this now because I'd add walls and light roof to third story and see how adding trusses can affect the load. My designers design the trusses don't worry. But they still keep telling me epoxy is the best repair because it is tested for 10,000 psi.. but when explaining to them the 10,000 psi only works when you compress it more than concrete (and strain incompatibility). They don't understand it anymore. I'm not making this up. But I can authorize them to approve any repair I want (such as manually removing each epoxy in cups and replacing with grouts) but the ground floor is already rented by tenants so can only do it if they leave.

 

[mes7a]

..

 

[mes7a]

http://www.eng-tips.com/viewthread.cfm?qid=395487

Kootk.. after learning more about interaction diagram and boundary conditions in columns from BAretired and how to derive the formulas. I think the safest is to retrofit braces frames, shear walls or even using stair shaft as lateral resisting system or turning it into gravity load only for the column with void at bottom.

Since you are familiar with my structure (before you forget them days or weeks from now). Please allow me to ask the following:

1. Since my bases are compromised (from voids filled with epoxy and insufficient flexural overstrength capacity in footing moment connections). The best way maybe to prevent rotations of the columns. Is it possible to brace the columns? with what? Like X frames between columns? How to put the X frames? Generally should such brace frame retrofit be connected to foundation.. it's difficult to dig the foundation again because of presence of tenants so shear wall may not be practical anymore.. I just want to prevent longitudinal movement (front to back.. see picture below) to avoid rotations of the columns that can stress the void in front (filled with epoxy) during extreme moment curvature). Of course I'll discuss this with my designers and they would design such system after getting some idea.. they don't have original ideas.

2. You mentioned above how stairwell shaft can be used as lateral resisting system. See:

In such system. Are the stairwell shaft connected to the columns or beams? In my layout.. it's not connected to middle columns so I wonder how you can consider it as main lateral resisting system.

3. You mentioned earlier that "In that case, the members not selected for primary lateral load resistance should be designed with sufficient deformation capacity that they can "ride along" with the designated lateral system without being compromised. I do this often.". For the column below:

Can it be made as column for gravity load only.. with the rest as moment resisting? But how do you make it have sufficient deformation capacity yet doesn't have lateral resisting effect.. you meant smaller column so it can't be converted (except by making the column smaller)?

Again. Thanks so much much. Don't worry we won't drag this thread to 200. Lol