Using Mud Slab to Resist Basement Wall Thrust 2

[KootK]

My situation shown below. This is making me pretty unpopular with the helical pile supplier and the only "outs" that I can think of are to:

1) Use the 2" mud slab like a regular SOG to resist lateral loads.

2) Use the backfill around the base of the wall to resist lateral loads.

Thus far, I've not been willing to do either of these things owing to concerns over of quality and permanence. How do others feel about this? Am I being too much of a hard ass?

 

[SJBombero]

So do the geotechs in your area also take responsibility for whatever structure may be sitting on top of that wall

If a structure falls down as a result of the poor recommendation of a geotech, they are responsible. Isn't that why they have a license? Isn't that why they are extremely conservative with their recommendations?

whatever structure may be sitting on top of that wall that is also experiencing seismic movement that may be imparted onto the wall at the same time?

I can begin to decipher what this means.

We are responsible for the seismic detailing of loads with overstrength as I noted above.

"Overstrength" is irrelevant to the topic at hand.

I don't see how them not providing it to the structural engineer that is designing the wall is within the letter of the IBC. We are responsible for the seismic detailing of loads with overstrength as I noted above.

Good geotechs will read all the relevant studies and come to the conclusion that for typical retaining walls <12' dynamic seismic force is zero. They have therefore complied with IBC. Less studious geotechs read a more narrow selection of studies whose applicability is limited and apply the conclusion of that study to everything. They also comply with IBC by recommended unnecessary dynamic forces. Both methods comply.
SEAOC Article 09.10.010 lays out adequate justification for why a geotechincal engineer may choose note to use an overly conservative abuse method to determine dynamic earth pressure. Again, just because the geotech came up with an answer of zero doesn't mean he didn't do the work to arrive at that answer. It is still and answer for which he is responsible.

 

[haynewp]

pvchabot,

I’m not going to spend any more time with it and overstrength is relevant. But you can keep doing what you’re doing and I will keep asking for seismic loads in high seismic areas from the geotech for the walls and the buildings supported on them that I am responsible for.

Here is some discussion on it for anyone else that is interested:

https://www.skghoshassociates.com/b...rth-pressure-on-basement-and-retaining-walls/

 

[SJBombero]

The current state of practice has been succinctly summarized in the attached paper:
Seismic Earth Pressures: Fact or Fiction

It appears that the current design practice for seismic earth pressures on building basement walls is conservative, uneconomical, and perhaps unnecessary. More importantly, the design practice is mostly based on experimental data that were extrapolated beyond the limits of their applicability.

https://files.engineering.com/getfi...4&file=Lewetal.-ER2010-ProceedingPaper066.pdf

 

[KootK]

Thanks for the continuing help gentlemen.

Updates based on conversations with the pile supplier:

1) It seems that we'll be using 1.5" square pile shafts rather than 3" pipe sections. This initially made me want to projectile vomit but I've mostly made my peace with it now.

2) As one might anticipate, those 1.5" shafts can only resist lateral loads via a single resistance mechanism: battering. So, wherever I can't mobilize passive/active soil resistance, which is most places in my opinion, my lateral loads will be resisted by battered piles. Batter batter batter.

3) The pile supplier's engineer took my question about the capacity design of the pile connection to his network and their interpretation is a bit different from what we've been contemplating. Rather than As x Fy, they take "short column" to mean a pile that is not governed by buckling but, rather, will have its connection force limited by the geotechnical capacity of the pile. This concerns me a bit in that, for capacity design purposes, I do not feel that I should be using the pile design capacity that I'm using for gravity loads and which is surely governed by settlement. I'd expect the true, load carrying capacity to be much higher. The piling engineer has mentioned torque values and how that establishes piling capacity. I'm having trouble understanding how that gets me to a true maximum capacity in a seismic, capacity design sense. That probably just reflect my own lack of familiarity with helical pile design. If anyone feels up to the task of enlightening me on this aspect of things, I'd be grateful to hear about it.

4) The piling engineer also suggested that just doing a basic, AScE overstrength design, rather than a true capacity design, might be a good fit here. I agree and may well take that approach if I can verify that doing so won't run me afoul of any code requirements.

 

[KootK]

Is your question "If I use battered piles, can I get out of the IBC connection provision?"

Yeah, pretty much. Thanks for you contributions to this so far.

My question to your question would be - what are you using the battered piles for? To resist overall sliding of the structure due to base shear or to resist out of plane loads due to lateral earth pressure? or Both?

Both although, with the greatly reduced out of plane loads, it is the in plane loads that are of primary concern now.

Are these piles battered in the plane of the wall?

There will certainly be some of that.

If so, I'd think that getting the connection to develop the strength of the short column would be fairly simple as it is just bearing on the concrete at that point.

Philosophically, I disagree. You could make that same argument about checking only bearing on the concrete at a conventional, column supporting pile cap. We don't do that though. Instead, we check punching shear. And, in the context of my situation, I see grade beam shear as the analog to that. I spoke to this a bit, previously, with this question:

2) If I have to apply this provision, how far do I take it? What qualifies as the "connection"? If it's just concrete bearing at the piles, no problem. I'd planned to extend this to include punching shear through the grade beams which is a bit trickier to deal with. Should I stop there? Consider some flexural stuff for the grade beams?

 

[KootK]

So, with these 1.5" piles, I now also have the more mundane concern shown below. Any thoughts on that?

 

[Enable]

I don't know much about helical piles or how they behave in/after seismic events so I am useless on the educational front. However, I can point you in the right direction as a professor at my Alma mater was pretty big into it. You'll want to take a gander at the work of Professor El Naggar. When I was there he was quite approachable and I'd recommend reaching out with your query.

You might find this bit interesting (seems like flimsy is best in these situations):

El Naggar and Abdelghany (2007a, 2007b) sought to quantify the amount that the ultimate axial compressive capacity of a helical pile reduces after a seismic event simplified by axial cyclic loads. They subjected three plain helical screw piles (HSP) and four grouted helical screw piles (GHSP) in cohesive soils to 15 load cycles of the same frequency, which was determined to be the average number of ‘effective’ load cycles of an earthquake. The ultimate capacity was determined using torque correlations and compared with the axial compressive failure loads of the final tests. Based on the results, it was concluded that after being subjected to 15 cyclic loadings the ultimate bearing capacity of the HSP decreased by 5%–10% and the GHSP by 18%.

Using observations from damaged pile exhumations from three earthquakes, Miura (1997) concluded that a ‘pile with higher rigidity will be damaged faster than a pile with lower rigidity when they are subjected to the same ground motion,’ and that ‘[a flexible pile] is better than a pile with higher rigidity.’ By all anecdotal success stories and case-history accounts of foundation performance under seismic load, flexibility due to the material property of the pile and possibly the looser soils within the zone of influence caused by installation disturbance, is one of helical piles’ greatest assets

The lateral deflection of piles is caused primarily due to the plastic deformation of soil. Even so, it has been found that helical piles recovered most of the deflection during unloading, indicating minimal structural damage (El Sharnouby and El Naggar 2011b).

 

[RFreund]

I have a soft spot for Helical Pier issues. I've taken a few deep dives into the subject.

Philosophically, I disagree. You could make that same argument about checking only bearing on the concrete at a conventional, column supporting pile cap. We don't do that though. Instead, we check punching shear. And, in the context of my situation, I see grade beam shear as the analog to that. I spoke to this a bit, previously, with this question:

OK, I've got it now. So yeah, I disagree with myself and agree with you. I was imagining a Chevron situation as you mentioned to avoid the shear situation. I'm also thinking that the shear wouldn't be that high but even with 1.5" square you're probably around 100kips, so yeah, that's high. But if you have a 36" deep grade beam, it might just be a few stirrups in that area.

Rather than As x Fy, they take "short column" to mean a pile that is not governed by buckling but, rather, will have its connection force limited by the geotechnical capacity of the pile. This concerns me a bit in that, for capacity design purposes, I do not feel that I should be using the pile design capacity that I'm using for gravity loads and which is surely governed by settlement. I'd expect the true, load carrying capacity to be much higher.

Hate to say it, but my first thought is that I agree with you here. I'm thinking of a very short load where the column is not buckling. The soil might "give" (i.e. settle), but it's not really failing. It's like asking, will the soil give enough or would that pile punch through the GB first?

There was another guy who was very helpful with a couple of my previous helical pier explorations. I'll see if I can find his contact info.

 

[phamENG]

Ah, torque values. A lot of helical pile suppliers will swear that's all you need. Through quite a bit of testing, they've determined a relationship between torque applied to the pile during installation and the load capacity. It's all empirical. And it's great...unless it isn't. Hit some debris? Torque value could hit the magic number. Hit a lens of stiff soil in a column of junk? Torque value could hit the magic number. So, for critical applications, it's important to have good boring logs and a geotechnical engineer to provide guidance. With that information, the geotech can estimate where the pile should be - so if you hit your torque rating 15' too short...you may stop and question it. The torque rating also does zero for settlement issues. You could hit a nice, 'competent' bearing strata...that just happens to sit over 12' of peat.

I think your concern about eccentricity induced bending is very valid, especially where your pile is cantilevering in the air. Maintain continuity where you can, design for eccentricity where you can't.

 

[RFreund]

So, with these 1.5" piles, I now also have the more mundane concern shown below. Any thoughts on that?

Is this separate than the foundation questions?
This feels like the "do you need to consider bending moments when a steel beam frames into a steel column with a simple connection" type of deal. I suppose you could take "e" as 1/2 the required bearing width and that would hopefully be small. There are also some tolerances that the piles are tested with to account for some eccentricity, but this doesn't seem to be that type of eccentricity.