Shear Wall Hardware Manufacturer Claims ACI 318 Anchorage Requirements Don't Apply 15

[D.E.N.]

TL;DR:
A shear wall hardware product manufacturer is claiming a steel threaded rod with epoxy anchored to concrete does not need to comply with ACI 318 anchorage provisions. I disagree. If ACI equations are used to check the anchor’s capacity, the product would have a reduced capacity and wouldn’t work as they have designed.

Full explanation:
I am the EOR on a large 5-story light-framed wood apartment building. I designed and detailed all of the shear walls using “typical” hardware, such as coil straps floor-to-floor and holdowns with threaded rods in epoxy at the base. Recently, I received a hardware substitution request from the GC, proposing a different product. Initially, I assumed this was a request to substitute hardware directly from a different manufacturer. However, the submittal proposes a completely different shear wall system.

The proposed system is a proprietary continuous hold-down assembly that connects to the building at the roof level only and anchors to the concrete at the bottom. Its tension capacity is based on testing and comes with a “code approval” report from a certified testing agency. This report, a Technical Evaluation Report (TER), is similar to an Evaluation Service Report (ESR) from other agencies. Due to the proprietary nature of the assembly, no published calculations or test result data are available; only a single allowable tension capacity is provided. The TER specifies installation conditions that must be followed to achieve the published tension capacity.

I have several concerns about this proposed system's ability to resist the required shear wall overturning forces and the proposed load path. Although I have been working with the manufacturer to address these concerns, one aspect of the design remains problematic. Both the manufacturer and the testing agency assert that the system's anchorage into the concrete does not need to be checked using ACI 318 anchorage provisions. The proposed shear wall system consists of a steel cable assembly with steel threaded rods at each end, with the bottom rod embedded in an epoxy-filled hole in the concrete. In my opinion, this anchorage design is not proprietary since it involves a steel threaded rod with epoxy in concrete, which is clearly defined in ACI 318. I would understand the argument if we were talking about a Simpson “LSTHD” Strap-Tie Holdown or MiTek “LSTAD” Foundation Strap, which is a truly proprietary anchorage design since it is just bent steel embedded in the concrete. But this is a steel threaded rod embedded in epoxy that just happens to be attached to the end of a proprietary cable assembly.

The manufacturer has a proprietary epoxy with an ESR report, stating to follow ACI 318-14 Chapter 17. However, the testing agency and manufacturer claim that the ESR for the epoxy is not valid when used with the cable hold-down system. They argue that the ESR only applies when the epoxy anchors steel threaded rods to concrete. It just so happens that the bottom of the cable hold-down system has a steel threaded rod attached to it…

Problem:
Using ACI 318 equations to check the anchorage capacity, I find that the assembly's allowable tension capacity is about 60% less than the manufacturer’s published values, primarily due to concrete breakout. This calculation includes product-specific variables and factors from ACI 355.4 tests, accounting for any proprietary epoxy behavior, which are published in the ESR.

I believe the proposed system’s anchorage must be checked using ACI 318 provisions, as the design closely matches the scenarios covered in the standard. The difference between the published assembly capacity and calculated anchorage capacity is a significant concern that I want to resolve before allowing the product to be used in my building. The testing agency and the manufacturer are telling me not to worry about it since the submittal and the TER are both sealed by PEs in the project state. They claim that since it is a delegated design, I don’t need to worry about how it interacts with my building. I would like to reject the entire system and stick with the original design, but the contractor really wants to use the cable system.

Question:
Have you ever dealt with something like this before? I don’t see how I can allow this anchorage “design” to bypass the ACI 318 requirements just because they are calling a standard anchorage method “proprietary.”

 

[driftLimiter]

D.E.N. - I can agree with you that having the load path is probably the better alternative. In my region we are high seismic and shear walls are often controlled by drift. So pretty much anything more than 3 stories is getting ATS systems. It really depends how moist the lumber is when it hits the site. Around here they are bringing it almost green.

Also I feel like your plan statement to nail after the roof is completed is at least some kind of nod towards the shrinkage.

Regarding the concrete breakout I'm curious about two things....

1. How did you get from allowable loads to LRFD strength per ACI? Like simpson hold downs they publish the allowable load which is mostly related to the wood structure.
2. Could anchor reinforcement, or supplementary reinforcement through the breakout cone be contributing to the test results?

 

[phamENG]

But you are right, the proposed system would likely do a much better job addressing building shrinkage than my original design would.

I disagree strongly with this statement. When using individual pieces of hardware at each level, each of them only has to deal with a little shrinkage - small enough that it's not a consideration locally. Globally, you might get slightly more flexibility, but it won't be significant. With the cable system, they're 'preloading' it. Great. They're preloading a cable swaged to a threaded rod set in epoxy and pulling down on a wood plate perp. to the grain. Does anyone here really think that will remain tight in the long run even without shrinkage much less somehow shorten almost an inch to retain some amount of preload after the shrinkage has taken place?

 

[XR250]

Does anyone here really think that will remain tight in the long run even without shrinkage much less somehow shorten almost an inch to retain some amount of preload after the shrinkage has taken place?

I most certainly do not

 

[D.E.N.]

driftLimiter - Thanks for the feedback. It's helpful to know when/where others are using ATS systems.

To address your questions about the concrete breakout:
1. Somewhere buried in their documentation I found a conversion factor of 1.62. I have seen this on other resources as well and I think it's logical. (0.9D+1.0W)/(0.6D+0.6W)= 1.62 (assuming 30% dead load and 70% wind load split). So taking the LRFD load combination divided by the ASD combination. Then multiply the ASD value by 1.62 to get to an LRFD level load. I often just take the ASD wind load divided by 0.6 (ASD seismic by 0.7) which is probably slightly more conservative.

2. No. I wish. See this screenshot from the TER. Footnote #1 says "no supplementary reinforcement". It also says "uncracked", that's a whole additional can of worms. And it says "in accordance with ASCE 19", which is titled "STRUCTURAL APPLICATIONS OF STEEL CABLES FOR BUILDINGS" and makes no mention at all about concrete or anchorage design...

 

[D.E.N.]

phamENG & XR250 - I agree with you. I don't see how that is supposed to work.

 

[driftLimiter]

While I agree and I'm no proponent of cables, Im not sure how the local effect is small enough that its not a consideration.

Strap elongation at allowable loads are around 0.1" inches. Your talking 1/8" per floor of shrinkage potentially that effectively doubles your hold down elongation.

Again, maybe the walls around these parts aren't normally drift controlled. Maybe they are proportioned so they are not drift controlled. But if 2x hold down deflection is not enough to consider the effect how much is???

 

[Aesur]

I recently saw this type of system being used on a project and had all the same concerns people here have, this was for a 5 over 1. The system had large double anchor rods on an embed plate at each end wall location with a lot of studs. I didn't check to see if they are all the way up, but I would assume so as I believe they need to be the same number all the way up based on how this system works. Considering many of these were near plumbing walls and switches, I expect they will be cut up a good bit. Long term I just don't see this system working as they anticipate. I asked the contractor how it compared in cost and they said it's about the same with all things considered, so the way I look at it is, cost is the same, but you need more wood, more anchors but less pieces compared to ATS just isn't worth it IMO. Other issues I have is, if you are preloading to 140% of the tension force required, is this additional loading being account for in the slab design? If this is a foundation versus podium, then does that mean you need larger footings now? I just don't see this making sense by any means and would just reject it.

I agree with driftLimiter here, anything over 3 stories goes ATS for us.

 

[driftLimiter]

2. No. I wish. See this screenshot from the TER. Footnote #1 says "no supplementary reinforcement". It also says "uncracked", that's a whole additional can of worms. And it says "in accordance with ASCE 19", which is titled "STRUCTURAL APPLICATIONS OF STEEL CABLES FOR BUILDINGS" and makes no mention at all about concrete or anchorage design...

Ya i got too curious after I asked and found this table and its footnotes. Its a hunk a junk if you ask me.

Did you treat the concrete as cracked or uncracked in your anchorage evaluation? Now i'm just really curious.

 

[Aesur]

It also says "uncracked", that's a whole additional can of worms.

Aren't all anchors in tension required to be "cracked" with no exceptions? in fact, I don't ever use "uncracked" as I just don't see how concrete can be considered that if under any kind of tension loading.

 

[driftLimiter]

@Aesur - I also always use cracked. The line of reasoning I have heard to demonstrate the concrete is uncracked is either: Anchor in the compression side of an element that is in obvious compression in perpetuity, or based on same hairbrained analysis of the cracking moment (just a guess if you ask me).

I don't think we are really talking about the anchor itself being in tension, but rather the concrete/rebar composite element.

 

[D.E.N.]

driftLimiter - I did my calculations as uncracked just for comparison purposes with their "allowable load" just because I know they would try to argue with me about the difference. I always consider cracked for my anchorage designs. Even assuming uncracked to match them, it was still about 60% off.


Aesur - I agree. I don't think you are allowed to consider uncracked for something like shear wall overturning anchorage. I considered cracked my original design.

 

[D.E.N.]

If anyone is curious, here is the calculation I sent to the manufacturer for them to review and tell me why there was a discrepancy. This uses ACI 318-14 Chapter 17 and the epoxy ESR.

 

[bones206]

And that’s without considering the sustained tension reduction factor, which is a drastic reduction. I don’t see that considered in any of their documentation, so in my mind it doesn’t meet code. At least for the adhesive anchor version. Again, I would not even entertain a post-installed anchor for this system.

 

[jerseyshore]

I've never used an ATS system (or even seen one), but I do know DrJ likes to pump out some shady stuff. They'll stamp anything for a check.

When it comes to new products I always require an ESR before approving. If they can't supply that I am not going to be the test dummy.

It is a bit surprising to see the response from the manufacturer saying something doesn't have to be designed per code. In the past, when I've contacted manufacturer's on certain products for more info they almost always give me the exact code section they used for design.

And honestly, if this was something that didn't have much research on it, or maybe direct code references, I might entertain it for more than 8 minutes.

But realistically, how many anchors does any one of us design in a given week? Dozens? Hundreds? Sometimes thousands. This isn't new technology that needs testing. There's probably a lot of people here that specify more anchors than concrete members.

 

[XR250]

Honestly, the more designs I do and the more time I spend reading ESR reports, the less I want to use proprietary products at all. A lot of these products seem to have inflated design values (based on testing or whatever), but if you read the fine print, it turns out the testing is really only for a very specific condition, which might be close to what you have for your building, but is probably not exactly the same. If something were to go wrong, I'm sure they (and their lawyers) would point to the fact that the actual conditions don't match the test conditions. A lot of these products look great on the surface but the fine print really makes it seem like they're trying to limit as much liability as possible.

Man, I feel the same way. I have never run into a proprietary product that did not have the fine print you are talking about. Even Simpson stuff can be sketchy at times.
I ran into this a bunch when I was doing CFS design.

 

[XR250]

Speaking of potentially sketchy proprietary products, what do y'all think about this one?

https://www.diamondpiers.com/#:~:te... Pier foundations,solar panels and much more.

I honestly do not understand how this was evaluated. They just base it off the manufacturers data? Has anyone actually done a full scale test?

 

[DoubleStud]

I always feel the way they test these anchors are sketchy. You are essentially pushing as much as pulling. The pushing force is essentially confining the concrete around the rebar. How is this a reliable test? The supports need to be far away from rebar or at least outside the ACI breakout cone. [URL unfurl="true"]https://res.cloudinary.com/engineering-com/image/upload/v1720963676/tips/IMG_0060_ncf4lo.webp[/url]

 

[Smoulder]

@Doublestud That's a test for bond. Basically whether it was installed properly. It doesn't test concrete break out because not the chem anchor manufacturer's or installer's problem. Also would need to test the whole group not individual anchors. Never seen any commentary on whether confinement in the test increases bond though.

 

[lexpatrie]

My problem is the "roof only" systems with wood will have such excessive shrinkage under normal construction details that it's just not going to even bear on the top of the wall.

It's possible that the ACI 318 anchorage provisions were not intended for shear walls, but the ACI 318 is for "not that redundant" connections, which a holddown surely FEELS like a not that redundant connection to me. I've specified a Simpson system (I think, or rather I tried to), and a Mitek Z3 was eventually used as their engineer submitted full calcs for the system. But it was into a cast slab on ground, full foundation, rather than a podium.

Epoxy anchors (to me) don't have the necessary capacity, I've gotten into a lot of disputes where the contractor decided they could dump the cast-in anchors and provide epoxy, it did not go all that well. One the original EOR has it as Exposure B and it was clearly Exposure C (hint: anything next to a Wal-mart is Exposure C unless proven otherwise). I got it to work because the main direction was loadbearing and the original design didn't consider that, but they didn't learn their lesson despite an overt warning letter, "Next time you decide not to install as specified, you're on your own. Because it probably can't be fixed without ripping out the concrete and starting over." Epoxy also requires special inspection from any epoxy product I know of.

 

[D.E.N.]

Update for anyone interested:
I told the GC that I could not approve the system as specified and that I needed the manufacturer to provide calculations (or any other proof) showing that their anchorage designs met the ACI 318 requirements. Due to the potential cost savings, the GC still wanted to use the system so they decided to put extra pressure on the manufacturer to provide the calculations I asked for. The GC just informed me that the manufacturer agreed that they would not be able to provide what I asked for to approve their system, so we will not be using it on this job. So we are back to hold-downs and coil straps (and ACI 318 compliance).